DEVELOPMENT OF SOUNDING EQUIPMENT FOR THE ASSESSMENT OF THE TIME-SETTLEMENT CHARACTERISTICS OF RECENT ALLUVIAL DEPOSITS WHEN SUBJECTED TO EMBANKMENT LOADS Geraint Alan Jones SUBMITTED IN PARTIAL FULFILMENT OF THE REQUIREMENTS FOR THE DEGREE OF DOCTOR OF PHILOSOPHY IN THE DEPARTMENT OF ENGINEERING , UNIVERSITY OF NATAL, 1992 11 ABSTRACf Many embankments on the soft, highly variable, recent alluvial deposits along the South African coast have suffered large settlements necessitating ongoing costly repairs. Due to the soft variable soils, borehole sampling is difficult and laboratory testing requires to be extensive for adequate subsoil modelling; cone penetration testing was considered to be a potential means to overcome these problems. Twenty five years ago in South Africa, as elsewhere, cone penetration testing equipment was relatively crude and the methods of interpretation were simplistic. The application of cone penetration testing to recent alluvial deposits therefore required improvements to both the equipment and the derivation of soil parameters. The equipment was upgraded by introducing strain gauge load cells capable of measuring cone pressures in soft clays with adequate accuracy. Hence, correlations of cone pressures with compressibility and shear strength became possible. Predictions of settlement times and magnitudes are of equal importance and a consolidometer-cone system was developed to assess both of these. A piezometer was incorporated into a cone to ascertain whether the settlements were due to consolidation. The piezometer cone performed so well that it superseded the consolidometer-cone and by 1977 a field piezometer cone was in regular use. Developments in piezocone interpretation have taken place concurrently with those in equipment; coefficients of consolidation are evaluated from pore pressure dissipations, and soils identified from the ratio of pore and cone pressures. These developments have been validated in two recent research projects, by comparing measured and predicted settlements at eleven embankments monitored for up to fifteen years. The data shows that for embankments on the recent alluvial deposits the constrained modulus coefficient, am is : am = 2,6 ± 0,6 The data also shows that coefficients of consolidation from piezometer cone dissipation tests are correlated with those from laboratory tests and back analysed embankment performance as follows : Embankment c = 3 CPTU c = 6 Lab c v v v It is concluded that piezometer cone penetration testing is particularly suitable for the geotechnical investigation and the subsequent design of embankments on recent alluvial deposits and should be considered as complementary to boreholes with sampling and laboratory testing. The existing database of embankment performance should be expanded with particular emphasis on long term measurements and on thorough initial determination of basic soil parameters. DEDICATION This work is dedicated to God Wendy My family ill PREFACE The whole of this thesis is my work unless specifically indicated to the contrary in the text, and has not been submitted in part or in whole to any other University. Some thirty years ago the author operated a deep sounding machine, one of the first in the country, on a misty lake in Ireland and marvelled at the way subsoil information could be garnered. The magic of the moment never entirely passed and when the opportunity arose to use the technique in Natal the die was cast. The development of the national road system surged in the early 1970's and since many of these roads on the Natal coastal routes crossed extensive recent alluvial deposits, the geotechnical problems of instability and settlement became major factors in the road design. Traditional methods of investigation consisted of boreholes with sampling and laboratory testing. Whilst these were satisfactory, provided they were of adequate quality, they were relatively expensive if sufficiently detailed models of the subsoil were to be obtained for design purposes. Cone penetration testing provided a potential a solution and this led to research work conducted over a period of twenty five years which continues today. The initial development of ideas for improvements to the mechanical equipment took place whilst the author was carrying out preliminary investigations for freeway routes over the coastal alluvial deposits. This was followed by a period devoted largely to cone penetration testing research and deVelopment and to embankment design methods at the National Institute for Transport and Road Research, and to the initial registration for a Master's degree under the supervision of Professor K Knight in 1975. This research programme was completed as originally envisaged, but not submitted because during its course the author conceived the idea of the piezometer cone. This proved to be such an exciting prospect that the research and development continued for a number of years until piezometer cone testing has now become almost routine for geotechnical investigations on alluvial deposits. In 1983, due to Professor Knight's retirement from the University, Mr Phillip Everitt was appointed as the supervisor. At that stage piezometer testing was becoming accepted internationally and new aspects and information frequently appeared. It was apparent, however, that the essential proof of the system for the prediction of embankment performance was to use it at IV embankments where the performance had been monitored. Eventually grants were provided by the Department of Transport for this, which enabled two research projects to be conducted during 1989 - 1990 and 1991 - 1992. After completion of the first of these a presentation of the author's work on cone penetration testing since the mid 1960's was made to the Faculty of Engineering at the University of Natal. The Executive Committee of the University Senate subsequently approved, in August 1991, that the registration be upgraded to doctoral status. Mr Everitt's encouragement during this extended period has been a vital factor in ensuring an outcome for this task and the author wishes to express his gratitude for this. v TABLE OF CONIENfS ABSTRACf DEDICATION PREFACE ACKNOWLEDGEMENfS PART A : INTRODUCfION, PROBLEM DEFINTI10N AND GEOLOGY Al INfRODUCTION A2 PROBLEM DEFINITION A2.l Embankment Engineering Problems A2.1.1 Stability A2.1.2 Settlement A2.2 Investigation and Analysis A2.2.1 Stability A2.2.2 Settlement A3 GEOLOGY AND CLIMATE A3.1 Geological History A3.2 Estuarine Deposits A3.3 Oimate PART B : IN1ERNATIONAL REVIEW OF CONE PENETRATION TEsTING Bl B2 B3 INfRODUCTION REVIEW OF MECHANICAL CONE PENETRATION TEST, 1930 - 1970 REVIEW OF PIEZOMETER CONE PENETRATION TESTING B4 CURRENT MEfHODS FOR TIlE INfERPREfATION OF CONE PENEfRATION TESTING B4.1 Introduction 11 ill 1 4 8 8 9 10 10 11 14 14 19 21 23 25 29 40 40 B5 VI B4.2 Shear Strength from Cone Penetration 40 B4.2.1 Cohesive soils 41 B4.2.2 Cohesionless soils 44 B4.3 Compressibility from Cone Penetration Testing 45 B4.4 Consolidation Characteristics from Cone Penetration Testing 59 B4.5 Soils Identification"from Penetration Testing 74 B4.5.1 B4.5.2 Mechanical friction sleeve soils identification Pierometer cone soils identification GEOTECHNICAL DESIGN OF EMBANKMENTS 74 76 87 PARTC: SOUTII AFRICAN DEVELOPMENTS IN CONE PENETRATION TESTING C1 lNTRODUcnON C2 SOUTH AFRICAN MECHANICAL CONE PENETRATION TESTING (1950 - 1975) 97 97 C3 MECHANICAL CPT EQUIPMENT AND INTERPREfATION DEVEWPMENfS C4 C5 C6 IN SOUTH AFRICA 111 C3.1 Methods of Estimating Embankment Settlements using CPT 111 C3.2 C3.3 C3.I.1 de Beer and Martens C3.I.2 Coefficient of compressibility, II\. Improvements to CPT Equipment - Vane Shear Improvements to CPT Equipment - Friction Ratio C3.4 Improvement in Interpretation of CPT Results - Friction Ratio CPT AS IN SITU CONSOLIDOMETER C4.1 Introduction C4.2 Research Project 1973 - 1976 C4.2.1 Prototype 1 (July 1973) C4.2.2 Prototype 2 (Dec 1974 - Jan 1975) C4.2.3 Analysis of cone settlement C4.2.4 Prototype 3 (May 1975 - June 1976) C4.3 Analysis and Discussion of Test Results LABORATORY PIEZOMETER CONE DEVEWPMENT OF SOUTH AFRICAN FIELD PIEZOMETER CONE 111 112 115 118 120 124 124 124 125 126 129 133 134 150 158 Vll PART D : SOUI1I AFRICAN APPliCATION OF PIEZOMElER CONE PENEfRATION JESTING D1 INTRODUCfION 172 D2 SITES DESCRIBED IN AUTHOR'S PAPERS (APPENDIX I) 172 D2.1 Umhlangane - Sea Cow Lake - (Jones, Ie Voyand McQueen, 1975) 172 D2.2 Mtwalumi (Jones, 1975) 173 D2.3 Uvusi (Jones, 1975) 173 D2.4 Umgababa (Jones and Rust, 1981) 173 D2.5 Urnzimbazi (Jones et aI, 1980) 174 D2.6 Sabi River, Zimbabwe (Rea and Jones, 1984) 174 D2.7 Waste Ash Dam - Kilbarchan (Jones and Rust, 1982) 175 D2.8 Tailings Dam - Bafokeng (Jones and Rust, 1982) 175 D3 DIVERSE APPliCATIONS OF CPTU 176 D3.1 Tailings Dams 176 D3.2 Gypsum Waste Dam 177 D3.3 Scour Depth - Umfolozi 177 D3.4 Natural Ground Level Identification - Richards Bay 178 D3.5 Irrigation Scheme Feasibility - Makatini Flats 178 D4 CPTU RESEARCH PROJECf, 1989-1990 179 D4.1 Introduction 179 D4.2 Back Analysis for Umgababa 181 D4.2.1 Drainage path length 181 D4.2.2 Loading 183 D4.2.3 Settlement record 184 D4.2.4 Rate of settlement 184 D4.2.5 Degree of consolidation 184 D4.2.6 Consolidation model 188 D4.3 Settlement and Time Settlement Predictions from CPTU at Umgababa 189 D4.3.1 Compressibility correlation 189 D4.3.2 Consolidation correlation 190 D4.4 Application of Umgababa Derived Parameters to Umzimbazi and Umhlangane 190 D4.4.1 Urnzimbazi 191 D4.4.2 Umhlangane 192 D4.5 Summary of Results 194 IX LIST OF TABLES B4.1 B4.2 B4.3 B4.4 B4.S B4.6 C3.1 C3.2 C4.1 C4.2 C4.3 C4.4 C4.S Coefficient of constrained modulus for normally consolidated and lightly overconsolidated clays and silts (after Sanglerat, 1972) Coefficients C1 and C; for different material types Modified time factors T· from consolidation analysis Subsoil description and coefficient of consolidation Bq values for different soils types Comparison of Bq values at soil type boundaries Material description from friction ratios Relationship between soil description, friction ratios, particle size, plasticity and coefficients of consolidation Particle size, Atterberg limits and cone t50 for consolidometer- cone tests Consolidometer and consolidometer-cone t50 and ~ times for Sea Cow Lake Consolidometer and consolidometer-cone t50 and ~ times for TSSH Modified consolidometer-cone t50 and ~ times for TSSH Comparison of cone / consolidometer ratios for t50 and ~ for TSSH and Sea Cow Lake C4.6 Consolidometer and consolidometer-cone times and ratios for all . C4.7 C4.8 C4.9 D4.1 DS.1 DS.2 DS.3 samples Measured strains for consolidometer-cone tests Consolidometer test results Cone : consolidometer settlement ratios Measured and predicted settlement data Constrained modulus coefficients, urn' from settlement analyses Constrained modulus coefficients, urn' from laboratory and CPTIJ data Coefficients of consolidation from laboratory, CPTIJ and settlement analyses 48 SO 73 76 79 84 114 122 138 140 142 143 144 146 147 147 148 19S 219 220 226 x LIST OF FIGURES A2.1 Topography and main roads of Natal 6 A2.2 Rivers of Natal 7 A3.l.a Geology of Natal 15 A3.l.b Geological Legend A3.l.e Lithology A3.2 Geology of Durban and environs 18 B2.1 Cone penetration test rig - 1930 25 B2.2 100 kN cone penetration test rig - 1960 26 B2.3 CPr rig with rods and penetrometer 28 B2.4 Mantle and friction sleeve cones 28 B2.5 Electric friction sleeve cone 28 B3.1 Piezometers tested (penman, 1961) 30 B3.2 Response times of piezometers (penman, 1961) 30 B3.3 Schematic of the piezometer probe (Wissa et ai, 1975) 31 B3.4 Time-rate of dissipation of excess pore pressures (Wissa et ai, 1975) 33 B3.5 a) Dissipation of pore pressure u with time (peignaud, 1979) b) Value of Uy as a function of OCR 35 B4.1 Undrained shear strengths in different tests (Wroth, 1984) 42 B4.2 Cone factors from strain path method (Houlsby and Teh, 1988) 43 B4.3 Variation of cone factor with !J. (Houlsby and Teh, 1988) 43 B4.4 Compression index Cc versus cone pressure, Rp (Gielly et ai, 1970) 48 B4.5 (1 + eJ/Cc against R,I OC (Gielly et ai, 1970) 48 B4.6 Constrained modulus coefficient versus cone pressure 51 B4.7 Water content, undrained shear strength and constrained modulus against effective overburden pressure (Coumoulos and Koryalos, 1977) 55 B4.8 Variation of strength ratio with plasticity for normally consolidated clays (Kenney, 1976) 56 B4.9 Constrained modulus against water content (Coumoulos and Koryalos, 1977) 56 B4.1O Plasticity index versus liquid limit (fsotsos, 1977) 58 B4.11 Compression index versus water content (fsotsos, 1977) 58 Xl B4.12 Comparison of predicted and measured coefficients of consolidation in Boston Blue Clay (Baligh and Levadoux, 1980) 63 B4.13 Dissipation curves for predicting '1t (probe) (Baligh and Levadoux, 1980) 66 B4.14 Pore pressure dissipation around spherical and cylindrical probes (fortensson, 19TI) 68 B4.15 a) Schematic features of consolidation around a cone b) Correlation between t50 and ko'p/Vw for Champlain Sea Clay (Roy et al, 1982) 69 B4.16 Values of'1t from laboratory tests (Sills et al, 1988) 71 B4.17 a) Excess pore pressure dissipation curves for different filter positions b) Excess pore pressure dissipation curves for modified time factor, T·. (Houlsby and Teh, 1988) 73 B4.18 Relationship between friction ratio and soil description 74 B4.19 Plasticity index against percentage <20J,Lm for Natal alluvial clays 74 B4.20 Soils identification chart (Jones and Rust, 1982) 78 B4.21 Pie:rocone net area correction 78 B4.22 Soils identification chart (Jones, 1992) 79 B4.23 Variation of pie:rocone pore pressure ratio with OCR at Onsoy (Wroth, 1988) 82 B4.24 Basis for estimation of OCR (Schmertmann, 1978) 83 B4.25 Classification chart based on cone resistance and pore pressure ratio (Senneset and Janbu, 1985) 83 B4.26 Soil behaviour type chart from CPTU data (Robertson et al, 1986) 85 BS.1 Settlement coefficient versus pore pressure coefficient (Skempton and Bjerrum, 1957) 88 BS.2 Terminology used for oedometer tests 89 BS.3 Relative importance of immediate settlement (Davis and Poulos, 1968) 91 BS.4 Error in settlement for one dimensional approach (Davis and Poulos, 1968) 91 BS.S Ratio of Eu/ Cu against OCR for clays 92 X11 BS.6 Relationship between settlement ratio and applied stress ratio for strip foundation on homogeneous isotropic elastic layer (D'Appolonia et al, 1971) BS.7 Relationship between initial shear stress ratio and OCR (Ladd et al, 1977) BS.8 Reduction of E with increasing stress level (Ladd et al, 1977) u C2.1 Cahbration chart and conversion table C2.2 Chart and table to derive 0 from'1cdfPb{vtxJ C2.3 a) Typical loose to medium dense silty and clayey sand b) Typical layered soft to finn clay and loose to medium dense sands C2.4 Dalbridge plate loading test (1965) C2.S a) Load settlement CUlVes for 20 ft. square concrete slab b) Progressive settlement of comers of a 20 ft. square concrete test slab under a uniform full load of 4,000 Ibs per square foot C2.6 Electromagnetic SPT trip hammer C3.1 CPT settlement estimation chart C3.2 Vane shear apparatus C3.3 Vane shear against cone pressure C3.4 CPT strain gauge load cell C3.S Chart recorder for vane and CPT measurements C4.1 Consolidometer-cone schematic C4.2 Consolidometer-cone apparatus: Prototype 1 C4.3 Consolidometer-cone apparatus: Prototype 2 C4.4 Laboratory consolidometer-cone time-settlement tests LPC series 1 C4.5 Laboratory consolidometer-cone time-settlement tests LPC series 1B C4.6 Laboratory consolidometer-cone time-settlement tests LPC C4.7 Laboratory consolidometer theoretical time-settlement CUlVes C4.8 Consolidometer-cone apparatus: Prototype 3 C4.9 Laboratory consolidometer-cone time-settlement tests for TSPC series 1 and 2 C4.1O Laboratory consolidometer-cone time-settlement tests for TSPC series 3 and 4A 93 94 94 99 99 101 103 104 108 116 117 118 119 119 124 125 127 131 131 132 132 133 135 135 Xlll C4.11 Laboratory consolidometer:-~ "'-FJr1~t4"~L' "' . 1 = 80 5Lf;~ID rO!>;'lance gauge ~-Jomct., '- 60 _ Jb Plus 340 It nylon lubsure oncrernenl } r: ' t ' I' I ~ I t--- 101020Ib/h2.P 'I,' - . ' I Ii, - .,. 11 - Il. - Pressu", ;ncremcnl 20 to 30 Ib.f;n' It- 1-+-t+I-I·tlirk-'.--H-H1 U r-V' 1. - - 1 -~ 401---t- . 1---.Y-V; 1-t-!~' 1- ' ! - ~ 40 II --+-. 'i 'JI 20-1- , r-X i~V-;! Ili-l-- 11!= 20 i 11'---J_r--!-tti'H! i~'lt-r----t o 002 0(1.;0 '102 05 1 2 5 10 20 50 100 0 002 OOSO'l 02 05 1 2 5 10 20 t min "· t min ~ lao J Hjll~ 7 fI'fTC~I3f',~~1 i'~jl' +]~L~'f 80 " D. -1'I1J j. ,- - . Ii _ 4a Tip C wilh SourdCf'l gau-;\-. l- _ 1_ /' l. ~ . 'b Plus 91lO It poty1h.,ne lu,,", 1 1.1 v,'jl- ei: 60 - Pl'{'Ssure inco-omr.nl r:V J ,''<- " 1:1- .. " 1- 101020 Ib,f;n.' !J ,'- .. , I"--i--t-++l Il. IJ I (- ,- i I 401- - " I jl- . lit-- ei: IJi , 1- I' Y- U. I ,1- 20 , ff - u. 117r- i ---l-H-t-HIH 1--~-++Hj:"':I--:7'71~ '11 n ! - I- - !I-;- o 002 005 01 02 O~ 1 2 5 10 20 50 100200 500 1000 t m~ Calculated values - Observed values .-. Figure B3.2 : Response times of piezometers (Penman, 1961) min 31 The only other paper in ESOPT I 1974, which made any other significant reference to the influence of pore pressures was that by J anbu and Senneset which discussed effective stress interpretation of in situ static penetration tests. They and Schmertmann were primarily concerned with the proper interpretation of cone penetration testing to obtain shear strength parameters. The other papers at ESOPT I dealing with what was then called Dutch sounding, static or quasi-static penetration testing were almost exclusively concerned with two aspects : the growing use of electrical cones, which had been given a great impetus by the offshore oil industry, and the development of correlations of cone and friction sleeve values with shear strength, compressibility and soil type. """' !:oOVC flll l OCI("'Ul - - . Figure B3.3 : Schematic of the piezometer probe (Wissa et a1, 1975) I· 32 In 1975, a number of publications appeared which described the measurement of pore pressures during cone penetration. The Raleigh ASCE Speciality Conference on the In-situ Measurement of Soil Properties, 1975, contained three papers referring to pore pressure sounding instruments or piezometer probes, viz by Torstensson by Wissa Martin and Garlanger, and by Massarsch, Broms and Sundquist. All three papers described piezometer probes in which only pore pressures were measured during penetration, ie no facility existed for measuring cone pressures. The Torstensson and Wissa probes were very similar to one another and also to that described by Penman. Figure B3.3 shows the Wissa probe which has a small cone angle, 20~ compared to the 60° conventional cone penetrometers, and has a small filter element at the tip. Massarch et al described a probe which had filter elements at a number of positions and demonstrated the significance of the position on the pore pressures recorded. All systems gave electrical outputs to chart recorder systems at the surface via cables threaded through the rods. Both Torstensson and Wissa et al noted that the excess pore pressures generated during probing could be measured, as could the rate of dissipation of pore pressures on stopping penetration. The observations made by Wissa et al were particularly interesting and some are shown in Figure B3.4. They commented "During penetration of the probe the soil surrounding the probe fails in undrained shear. As a result, dense or stiff soils generate negative excess pore pressures when penetrated by ~he cone, whereas loose or soft soils develop positive excess pressures. The rate of excess pore water pressure dissipation with time after pushing of the probe is stopped is a function of the permeability and compressibility of the soil surrounding the tip of the probe. It should be possible to develop theoretical relations between these soil properties and the time rate of dissipation of excess pore pressure; nevertheless, until such relations are available, the data gives a qualitative indication of the type of soil being penetrated." In 1975 Richards et al described a piezometer probe which measured differential pressures in the sea floor primarily for the purpose of obtaining a better understanding of submarine slope stability. The opening statement of the paper is quoted verbatim: "More than a decade ago it was proposed that the likely relationship between excess pore water pressure in cohesive sediments and potential submarine slides, slumps and possibly turbidity current could be investigated by the in situ measurement of pore 33 pressure in sea floor sediments (Richards, 1962). Negotiations with the Norwegian Geotechnical Institute (NGI) in 1964-5 led to the acquisition by the University of Illinois, Urbana, of two differential piezometer probes and a counter computer in 1966. This paper describes the probe system that was built around these units and the results of its limited testing at sea in 1967. II: ... i ... o .. II: ... ... ... ::II 01 05 1.0 50 100 500 1000 10000 (LAPS(O TI"( IN MINUT(S - LOG SCALE Figure B3.4 : Time-rate of dissipation of excess pore pressures (Wissa et aJ, 1975) Previously the system was mentioned by Richards (1968) and Richards and Keller (1968). One test was very briefly summarised by Lai et al (1968)". It is surprising that this earlier work should not have evoked more response from those involved in cone penetration testing. The earlier references were to discussion at a Geotechnical Conference, Oslo, to a paper in the American Association Petroleum Geology Bulletin and in American Geophysics Union Transactions and these were either not seen or the implications not appreciated by other cone practitioners. To 34 some extent this demonstrates the difficulty in assessing claims for originality in the development of new ideas and equipment. In 1975, Levillain described a piezometer cone, la sonde piezometrique, which measured only pore pressures. The primary purpose was to emphasize the pneumatic pressure measuring system which could be installed in a number of different piezometers ie not necessarily only in the cone push in type. No results of testing in soils were given. In 1976 Parez, Bachelier and Sechet described the equipment and results obtained for a piezometer cone again only for measuring pore pressures not cone pressures. They used both a cylindrical filter in the shaft of the cone and individual small filters in the face of the cone and noted that very similar results in pore pressure magnitude and sign were obtained for both cones. They placed strong emphasis on the necessity for a rapid response instrument. The primary purpose of the work reported was to assess effective stresses around cones and hence adjust both cone and friction sleeve resistances to be used in the design of piles. In 1978 Schmertmann published a comprehensive report on the use of the Wissa 1975 type piezometer probe to identify liquefaction potential of saturated fine sands. He compared the performance of a Wissa 20~ 1,75 in dia cone and a Wissa 60~ 1,5 in dia cone, the latter being similar in shape and size to the standard Dutch cone, except for the filter element at the tip, and concluded that the probes gave similar results but with the 20° cone giving better layering detail and the 60° cone giving a greater magnitude of generated pore pressure. He mentioned that the rate of dissipation of pore pressure could perhaps form the basis of estimating permeability and described the de-airing procedures and precautions required to keep the filters saturated in the field. In 1978, Baligh, Vivatrat and Ladd published a comprehensive research document on the exploration and evaluation of engineering properties for foundation design of offshore structures which concentrated on cone penetration testing. This, inter alia, described further testing with the Wissa-type piezometer cone with the filter at different positions. It also summarized current theories on cone penetration. It noted that pore pressures dissipated on stopping penetration and that the study of dissipation rates and hence consolidation characteristics was being undertaken separately. 35 In 1979, Peignaud, described the driving of a cone shaped piezometer into clayey soils. He discussed the influence of the filter position on the pore pressures recorded and noted how this makes it difficult to compare test results using different cones. The tests were carried out at different penetration rates which was the primary emphasis of the work. In the clays tested the pore pressures generated were negative during penetration and became positive after dissipation to values dependent on the overburden pressure. At the range of penetration rates tested ie 5, 10 and 25 mm/s the negative excess pore pressures generated were penetration rate dependent but the final pressures were very similar for all rates. Peignaud showed an interesting result of negative excess pore pressure generated against degree of overconsolidation which is reproduced in Figure B3.5. Figure B3.5 : UkPa (a) 400 300 200 100 0 5 10 t.mp5 In min Uv kPo (b) - 300 -200 . -100 2 3 4 5 6 7 k O""vo a) Dissipation of pore pressure u with time (Peignaud, 1979) b ) Value of lly as a function of OCR In 1980, Baligh and Levadoux; Levadoux and Baligh; and Baligh, Azzouz and Martin published research reports on pore pressure dissipation after cone penetration, on pore pressures during cone penetration in clays and cone penetration tests offshore the Venezuelan coast. These detailed reports followed up the 1978 reported work. The 36 cones used were the Wissa type, both 18° and 60~ and with the filter elements in different positions, and had not changed since the 1975 versions other than in some relatively minor details. Also in 1980, Baligh, Vivatrat and Ladd presented a paper on soil profiling using pore pressure ratios which was a summary of the work described in their 1978 research report. From 1981 onwards, in the space of two or three years, piezometer cone penetration testing measuring both cone and pore pressures moved forward in almost a quantum leap from an experimental tool to an off the shelf commercial production unit. The author was responsible for four papers in this period viz, for 10th Int. Conf. SMFE, Stockholm 1981; ASCE Symposium Cone Penetration Testing and Experience, St Louis, 1981; ESOPT II Amsterdam, 1982; International Symp. In situ Testing, Paris 1983. These described the South African development of and experience with the new piezometer cone which for the first time measured cone and pore pressures simultaneously. As far as the author is aware these papers, and others at the same conferences, were the first to describe such piezometer cones. The author's papers, as well as giving results for a number of sites where piezometer cone testing had been carried out, also gave a method of deriving coefficients of consolidation from pore pressure dissipations and a method for soils identification from a comparison of generated pore pressures with cone pressures. At the first of these conferences two other papers on similar piezometer cones appeared by Franklin and Cooper (1981) and by Parez and Bachelier (1981). Franklin and Cooper referred to their cone as a prototype developed from the Wissa probe in so far as the pore pressure measurements were concerned and having the filter element at the tip. The cone included strain gauge load cells for both cone and friction sleeve measurements. They did not draw firm conclusions from the limited testing in recent sands, but confirmed general trends previously observed by others using the Wissa probe. The paper indicated that the probe itself performed without problems provided that de-airing was carefully done and subsequent saturation ensured. 37 Parez and Bachelier described a piezometer cone with a conventional cone pressure measuring system and a cylindrical pore pressure measuring system in the shaft above the cone, ie strictly not a piezometer cone but having some of the same purposes. On the basis of the simplified approach of a cylindrical dissipation model used for sand drain design they suggested a time factor could be derived, for the estimation of coefficients of radial consolidation. At the ASCE Symposium on Cone Penetration Testing and Experience, 1981, there were a number of papers describing piezometer cones and testing viz Baligh et al; Battaglio et al; Campanella and Robertson; de Ruiter; Marr; Muromachi; Tumay et al; and that by the author and colleague, Jones and Rust. At the Second European Symposium on Penetration Testing, 1982, ESOPT II, some of the same authors had further publications together with papers from Abelev; Lacasse and Lunne; Marsland and Quarterman; Rocha Filho; Senneset et al; Smits; Sugawara et al; Tavenas et al; Torstensson; Tumay et al; Zuidberg et al including the author and colleague, Jones and Rust, (1982). Of the 86 papers on cone penetration testing 16 described piezometer cones and gave test results. In 1982 two papers were published in the Canadian Geotechnical Journal by Roy, Tremblay, Tavenas and La Rochelle which described a piezometer cone and its use in sensitive clays, and in 1983 by Campanella, Robertson and Gillespie on testing in deltaic soils. Later in the same year Robertson and Campanella published a paper in two parts on interpretation of cone penetration tests in sands and in clays. These were in essence state of the art papers based on world wide experience recorded in the referenced literature. In the following few years relatively little was published on piezometer cone testing that made any significant advances in the equipment or test methods, but the theoretical understanding had improved as better insight into the soil behaviour around a penetrating cone could be obtained from many more detailed field observations coupled with laboratory testing. 38 At the 11th ICSMFE, San Francisco, 1985, Campanella et al and lamiolkowski et al presented recent developments of in situ testing and new developments in field and laboratory testing which indicated little new development since ESOPT II in 1982. At the ASCE Speciality Conference, In situ 86, Blacksburg, a number of papers were published on piezocone testing by Keaveny and Mitchell; Lunne et al; Mayne; Olsen and Farr; Robertson et al. At IS OPT I in 1988 at Orlando, Campanella and Robertson presented a special lecture on the current status of the piezocone test. Essentially this had little new to report since the burst of information in 1981/82, but it reflected a much broader acceptance of the method and the search for greater understanding of the results obtained. At the instigation of those most involved in piezocone testing, including the author, a section was included in the International Reference Test Procedures on the piezocone with the intention of reinforcing acceptance of the technique. Special lectures were also given on cone penetration testing by Parkin, Mitchell, lamiolkowski et aI, Hughes, and Wroth. Wroth's contribution should be singled out for it urged that the many correlations suggested for the CPT results and in particular for CPTU results should "be based on physical insight, set against a theoretical background, and be expressed in suitably dimensionless form". At ISOPT I there were nineteen technical papers on piezometer cone testing out of a total of sixty CPT papers. Other than those describing special purpose cones, the remaining papers barely described the cones themselves; they concentrated on the results obtained from actual testing and derived relationships for defining overconsolidation, shear strength parameters, consolidation parameters and soil classification. Four of these discussed overconsolidation or stress history viz Mayne and Bacchus; Rad and Lunne; Sugawara; Sully, Campanella and Robertson and the importance of the OCR in predicting the behaviour of soils was stressed. The paper by Houlsby and Teh who used a strain path method and a large strain finite element analysis, has added significantly to the insights on cone penetration testing. A finite difference method is used for the dissipation analysis and a new method of interpreting piezocone pore pressure dissipation data is given. 39 Four of the papers at IS OPT I described the use of pore pressure dissipation tests in order to derive coefficients of consolidation viz Lutenegger et al; Senesset et al; Sills et al; Tang Shidong and Zhu Ziao-Lin. The methods used are all similar and could almost be called conventional by this stage. The paper by Sills et a1 is particularly valuable in that it compared coefficients of consolidation derived from the piezocone (using Baligh and Levadoux, 1980), from laboratory tests and from back analysis of the settlement of the embankment. The Lutenegger et al paper also compared piezocone derived parameters (using Gupta and Davidson, 1986) with laboratory and field data. They claim "excellent agreement" but in fact their results appear to show almost an order of magnitude difference between the field measured coefficients of consolidation (based on in situ piezometer readings) and the piezocone data, with the laboratory data spread covering this range. Senneset et aI, using Torstensson's spherical expanding cavity solution and their definition of the effective sphere radius and also the 1982, Senneset et aI, method, derived coefficients of consolidation, from cone dissipation tests and compared these to laboratory test results and claim the "data seem to correlate fairly well". The Tang Shi-dong paper, which is primarily concerned with piling, made only general reference to the dissipation of pore pressures insofar as they influence pile bearing capacity. The remaining dozen papers on piezocone in the 1988 ISOPT I discussed a variety of aspects which, although of interest to piezocone practitioners, are not directly relevant to this present work on embankment settlement. ISOPT I can be seen as representing the coming of age of piezocone equipment development and to a lesser extent of the interpretation methods. It was perceived that the initial almost dramatic advances had been made and that a period requiring solid definitive testing for data acquisition was necessary so that the most appropriate theoretical models could be calibrated. This is not an easy task since it will require high quality piezocone testing, high quality sampling and laboratory testing, followed by performance monitoring of constructed works. 40 B4.1. Introduction At the beginning of the period covered by the author's close involvement with cone penetration testing ie 1965, the cone systems on a world wide basis were the simple mechanical type. The Dutch 60°, 10 cm2 mantle cone, with or without a friction sleeve was the most popular of these. The body of knowledge was firmly in western Europe whence the methods of interpretation emanated. These were comprehensively described by Sanglerat (1972). At the earlier stages the emphasis of cone penetration testing had been on the assessment of the characteristics of cohesionless soils for the purpose of foundation design, primarily for piles. Certainly Sanglerat covered a much wider scope than solely pile design, but many of the contributions in the chapters describing experiences in different countries concentrated on this aspect and often on deriving correlations between what was then called Dutch Cone Penetration Testing, or quasi state penetration testing and Standard Penetration Testing - SPT - so that any design method applicable for the latter could be simply adapted for use with the CPT. Nevertheless interpretation methods for all aspects of the CPT were described by Sanglerat and have not in most cases significantly changed except insofar as the advent of the piezocone has allowed development of these methods. The following sections describe the current generally accepted interpretation methods as they apply to the establishment of the characteristics of alluvial deposits which are required for the design of embankments. These characteristics are shear strength, compressibility, consolidation, soils identification and over consolidation ratio. The application of these to embankment design is discussed in B5. B4.2 Shear Strength from Cone Penetration Shear strength derivations view cone penetration as a classic bearing capacity problem which can be expressed as : == B4.1 where q! cu °vo B Y 41 = = = = = ultimate bearing pressure undrained shear strength overburden pressure foundation width subsoil density and N", N q and N. y are bearing capacity factors depending on the shape of the foundation and the friction angle 4> of the soil. B4.2.1 Cohesive soils For the case of undrained shear in a clay where 4> = 0, since Ny = 0 when 4> = 0 and Nq = 1 then the equation B4.1 reduces to : = B4.2 where = cone pressure However, contrary to conventional bearing capacity theory which dealt with surface or shallow foundation and for which values of Nc could be computed and depended only on the geometry, it was shown by Gibson (1950) that for a deep circular foundation Nc was also dependent on the rigidity index Ir of the soil, where Ir = G/su: B4.3 At a typical rigidity index of 200 the equation B4.3 gives an Nc value of 9.4 and would require a rigidity index of about 104 to give an Nc value of about 15. Experience shows that this analysis for a deep circular foundation does not give appropriate values for Nc for cone penetration testing and the problem becomes one of determining such values. The well established soil mechanics route has been followed of empirically determining values of Nc by carrying out independent measurements of undrained shear strength, say from undisturbed sampling and undrained triaxial testing. Much research has been carried out in this way to determine Nc for a range of soils and indeed it continues to 42 be recommended practice that local correlations should be made for any clays where values are not already available. Tables of values are available based on descriptions of the clay type, the sensitivity and overconsolidation ratio. An ongoing difficulty with this approach is the implication that for a particular clay the undrained shear strength has a unique value. It has been generally recognized and discussed by Wroth (1984) that the undrained shear strength is a function of the type of test. Figure B4.1 taken from Wroth demonstrates this and it follows that to be consistent a hierarchy of Nc values should be used in equation B4.2 depending on which equivalent undrained shear strength is required from cone penetration testing. An alternative approach is to achieve consensus on how Nc should be derived for cone penetration testing whether it be strictly theoretical or empirical, and use this defined set of values to determine Cu (cone) which will then become an additional undrained shear strength in the existing hierarchy. Wroth (1988) recommends the former approach and suggests that, "undrained triaxial compression test (after appropriate reconsolidation of the specimen) - is used exclusively in all such correlations." Uncuamed shear strenglh: kPa or° -,~·r°-.~8r°-,~120 8 ~ 16 .c 0. ., c 32 (al CKoUC Triaxial compression CKoUE Triaxiar eX1enSlon DSS Direct simple shear FV Field vane Undr~ined shear strength: kPa o 0 .0 80 CKoUC (bl Figure B4.1 : Undrained shear strengths in different tests (Wroth, 1984) Houlsby and Teh suggest that the cone factor Nkt defined as : 43 can be represented as : 4 ( ~ 1 Nk;t = "3 (1 + InI,) 1,25 + 2000 + 2,4 ", - 0,2 "s - 1,Sa B4.4 where the last term 11 = (CJ' vo - CJ' ho) /2su is the initial shear stress ratio and is a correction to allow for the "incorrect" use of (qt - CJvo)/su instead of (qc - CJho)/su to define Nkt and the a terms relate to the cone face and shaft roughness. The second term with Ir is a correction since finite element modelling showed some deviation from the theoretical spherical expanding cavity solution. Figure B4.2 and B4.3 show Nkt varying with 11 and Ir when a is constant and demonstrates the high dependence on 11 and Ir. In practice the variation is conditioned by the fact that for any given soil, variations in 11 and Ir with OCR tend to cancel each other out. 16 Nkt Strain path method 12 kc 8 I. Spherical cavity expansion Cylindrical cavity expansion 0 0 50 100 150 200 I, Figure B4.2 : Cone factors from strain path method (Houlsby and Teh, 1988) Nkt 7 Figure B43 : Variation of cone factor with !J. (Houlsby and Teh, 1988) The Nkt values obtained agree more closely with those found in practice than any other theoretical method and this approach appears most promising. 44 A consequence is that equation B4.2 continues be valid but that the value of Nc can be derived theoretically and is dependent inter alia on the rigidity index. Unfortunately, however, this index has a very wide range and can only be measured by extensive laboratory testing. It must be expected, therefore, that values of Nkt (which have a fairly small range) derived from experience will continue to be used in practice albeit with more effort being expended to justify the Houlsby and Teh approach. B4.2.2 Cohesionless soils The author's work is concerned primarily with the application of the CPT to embankment design on predominantly cohesive soils so the shear strength of sands is relatively unimportant since it is not critical in embankment performance. Little discussion is therefore given to this aspect. For cohesionless soils equation B4.1 reduces to : = B4.S and Nq varies non linearly in the range 0 - SOO as 0 increases from zero to SO~ This suggests that qc should increase indefinitely as the depth increases but practice shows this does not happen and different materials have well defined upper bound values of qc- The simplistic model is therefore inadequate to cover all situations but nevertheless considerable experience has shown that cone resistance can be related to stress level, relative density and angle of shearing resistance. Many analyses have derived angles of friction, 0, directly from relationships between 0 and qc eg de Beer (1948), Harr (1977), Durgunoglu and Mitchell (197S), and it is probably the last of these which is favoured . The stress level has generally been represented by the ambient vertical effective stress a / vo rather than by the horizontal effective stress, a'ho' since a /vo can be calculated or estimated from the soil unit weight and hydrostatic conditions whereas a/ho cannot readily be obtained. Recent work, however, has confirmed that it is the horizontal stress that controls penetration resistance. Wroth (1988) categorically states that the conventional practice of normalizing cone pressure with respect to a ' should be vo discontinued and that a/ho' the horizontal effective stress should be used. Been and 45 Jefferies (1985) and Been et al (1987) have defined a state parameter which describes the condition of a sand. This parameter 1£1 is the difference between the actual void ratio of the sand, eo' and the void ratio, ess' at the steady state at the same mean effective pressure, p', (p' = 1/3 (a /vo + 2 a/ho)' The state parameter combines the influence of relative density and stress level and leads to the proposed relationship : 0,7 MN/m2 2 to 5 2.5 to 6.3 Silts of intermediate or low MI, ML 3 to 6 3.5 to 7.5 plasticity Organic silts OL 2 to 8 2.5 to 10 Peat and organic clay : Pt, OH 50 < w < 100% 1.5 to 4.0 1.9 to 5.0 100 < w < 200% 1.0 to 1.5 1.25 to 1.9 w > 200% 0.4 to 1.0 0.5 to 1.25 The results given by Gielly et al (1969) were also given in the form of Cc against qc and are shown in Figure B4.4. i . \ , 1 .: .·\ I ' . . . ; .. . \ • =: " : : 1 u~--;-_-;-_-;;-----;:_--:-_...". .. _--','-. _...J .. ...... ........ .......... ... Figure B4.4 : Compression index Co versus cone pressure, ~ (Gielly et a1, 1969) ] 0 '. 10 10 ]0 ]0 Figure B4.5 : (1 + eJ/Cc against ~/ (J c (Gielly et al, 1969) R. r;- 49 The data extracted from this comprehensive laboratory testing included Atterberg Limits and natural moisture contents and Gielly et al state that general agreement is shown with Skempton's (1944) equation for undisturbed normally consolidated clays although there is a large scatter: Cc = 0,007 (wL - 10) B4.12 The generally held view is that correlations of Cc with liquid limit are relatively poor so the scatter does not necessarily cast doubt on the Cc against cone pressure relationships and therefore on the "0 coefficients. Gielly et al also show a diagram of (1 + ec)/Cc against ~/oc where ec and 0c are the void ratio and consolidation pressure. (~is the probe resistance or cone pressure, qc). This is the straight line relationship, equation B4.13, shown in Figure B4.5, taken from Gielly et al paper. This serves to confirm that compressibility is proportional to the cone pressure but that the relationship is conditioned by the void ratio: 1 + e c B4.13 Schultze and Menzenbach (1961) and later Schultze and Mezler (1965) carried out laboratory and field tests to obtain correlations of SPT N values (1961) and subsequently cone pressures, qc' (1965), against coefficient of compressibility, 0\. They proposed equation B4.14 given below in which the constants C1 and ~ have a range of values dependent on the position of the field test relative to the water table and the nature of the material: B4.14 50 Table B4.2 gives values for C1 and C2 and it can be seen from the correlation coefficient that for sands and clayey sand the correlation is excellent and good respectively, but for sandy clay is much poorer. Table B42 : Coefficients C1 and C; for different material types Above Below Sand Clayey Sandy Loose Soil Type Water Water Sand Clay Sand Table Table No of tests 15 17 14 19 27 18 C1 (bar) 52 71 39 43 38 24 C; 3.3 4.9 4.5 11.8 10.5 5.3 (bar/blow) Correlation coefficient 0.758 0.900 0.954 0.886 0.783 0.764 Based on the Schultze and Menzenbach (1961) publication the author developed similar equations for alluvial deposits in Durban from the results of a large (6,1 m x 6,1 m) plate loading test and from screw plate tests. The equation B4.14 and factors C1 and C2 were adapted from SPT to CPT by using the then generally accepted qc to N value correlations and adjusted on the basis of the plate test results. These were initially reported by Kantey (1965) in discussion at Montreal and later in a paper by Webb (1969). These equations, devised largely by the author, are given below: Fine to medium sand : :::; 5/2 (qc + 3000) B4.15 Clayey sands (~ < 15) : 1 M :::; - :::; 5/3 (q + 1500) m c v B4.16 51 Estuarine deposits, mixed layered sands, silts, clayey sands, clays : M = 1 = 2 ( , .~. 0 :r,.,~ 0 . '-'f.:.' .~~ 0 .. '~ .. ~~ 0 ~-oo 'j-, I' "()ooloo ': .. ~ ~ . , , I Vcy/ Po~ o.09 , I ~\ ' i , \' r' \ .. 1 :, \ , , , \1 , 1 I , ;\ , \. , .. '1\: , "\ .. -I ' . , 1\' , r\ . 1 \ 1 , 200 0) b) c) Figure B4.7 : Water content, undrained shear strength and constrained modulus against effective overburden pressure (Coumoulos and Koryalos, 1977) Until such time as Eu/su can be comprehensively modelled it must be obtained empirically for specific soils. For example eoumoulos and Koryalos (1977) give data for soft normally consolidated clays near Athens expressed as shear strength and deformation modulus against effective overburden pressure, Figure B4.7. They state that the individually calculated results give a Eu/su in the range 100 to 170 and that the relationship is linear within small increments of stress. However if the linear regression lines of Su and Eu against effective stress are compared then clearly a linear relationship between the two is forced, with, in this case, a value of 280. Also the correlation coefficients for straight line modelling of both Su and Eu are low hence not overmuch significance should be given to the tentative Eu/su relationships shown, other than that they demonstrate a valid practical approach. No specific data is given in the paper on possible overconsolidation hence the statement that the clay is only about 3 000 years old and "can be considered as normally consolidated". However the cu/Po' relationship, which is a straight line as would be expected for a normally consolidated clay, shows an intercept with the shear strength axis at zero effective overburden pressure of about 56 15 kPa, which is higher than would usually be expected for a normally consolidated clay (up to 5 kPa). The author suggests therefore, that the upper clay may be lightly overconsolidated, and hence the low value of sui 0vo of 0,09 is not appropriate, which in turn invalidates the actual derived Ejsu value. The sui 0' vo can be compared with that reported by Kenney (1976) shown in Figure B4.8, or with the relationship between sui 0' vo and plasticity index values given by Skempton (1954), see equation B4.52. 0 0,8 .> b o ARTIFICIAL CLAYS , BJERRUM & ROSENQVIST ::J • FRESH - WATER 119561 ell 0 0,6 5 II: x I- " 0, Z 1&1 III: l- V) II: ~ 0, x CANADIAN , V) Q VARVED CLAYS 1&1 LEACHED 21 ~ II: a D z 0 20 ::;) 40 60 80 ' loa PLASTICITY INDEX, "10 Figure B4.8 : Variation of strength ratio with plasticity for normally consolidated clays (Kenney, 1976) on 3 20 :::;) o 0"1 ~ E 10 1 - 10 I ~ • r--:... • • • -. ~ • ~ • • "- 20 30 40 50 70 WAT ER CONT ENT •• ,. Figure B4.9 : Constrained modulus against water content (Coumoulos and Koryalos, 19TI) 57 From the Coumoulos and Koryalos results it is clear that the selected su/po' of 0,09 is not consistent with the above empirical relationship and should not be considered as representing a normally consolidated clay. Similarly it can be seen from their natural moisture contents and plasticity data that the liquidity index is generally low (approximately 0,5) and does not decrease with depth, as would be expected for a normally consolidated clay. The plasticity indices for the layer under consideration encompass a very large range of about 10 to 50 and it must be concluded that describing this layer as uniform is misleading. Not surprisingly the relationships of E/su and Su with effective overburden pressure and with one another are ill defined. This is not intended as criticism but merely to highlight the very considerable difficulties in obtaining reliable data which could be used not only on a local basis but transferred to a data base from which to build general relationships for the determination of Ejsu' Coumolos and Koryalos also give a relationship between constrained modulus and water content shown in Figure B4.9 which has a coefficient of correlation of 0,81. Whilst correlation with moisture content may be valid for a uniform clay, since the uniformity will by definition have eliminated most important characteristics - viz plasticity, stress state and overconsolidation, - correlation with a wider range of descriptors than solely moisture content must be necessary to generalize the situation. Tsotsos (1977) makes exactly this point and Figures B4.1O and B4.11 illustrate it. The lines shown on Figure B4.10 correspond to soils which fit the equation: B4.25 where A varies from 0,78 to 0,36 and B correspondingly varies from 4 to 35, the centre of the five lines being the standard "A" line with values of 0,73 and 20. Figure B4.11 shows compression indices against moisture content for the same soils (up to a liquid limit of 80) and demonstrates the influence of the plasticity characteristics on the compression index since the moisture content is dependent on the plasticity in saturated soils. Using this approach and obtaining similar data for local soils it is then possible to assess the change in compressibility corresponding to changes in natural moisture content and in this way the constrained modulus coefficient, am. could be modified on the basis of the natural moisture content. . . 58 \ 00 ~ qgo ] o,f!IJ I: .~ ~ qiU ~ ~ OO~~~~-+-4--~~ ~ ~~~+-4-~-+­ '0 Q 0.60 ~ V E ~~+-4-~-+~~~~~~~ 0.50 > .. 'u 30 ~+-4-~---\"...:.-j~,J..£--+- Q,40 .. ~ 20 ~+--+-~":-'4---,4~~~ Q. 0.20 O~~~=C~~ __ L-J--L~~L-~ 0;0 o 10 20 30 40 50 60 70 80 90 100 110 120 Water Content W·'. Liquid limit WL 0/. Figure B4.10 : Plasticity index versus liquid limit (Tsotsos, 19TI) o 10 20 30 40 50 60 70 80 90 100 110 Figure B4.11 : Compression index versus water content (Tsotsos, 1977) In summary the justification for the use of cone penetration testing in clays to measure compressibility remains that of practicality. The wealth of field data reported in the literature gives moderately consistent values of the constrained modulus coefficient, am' directly related to cone pressures, qc' The range of am values is, however, large and the selection of the appropriate value within the range is ill defined which results in crude estimates of settlement. It appears that better definition of the appropriate am value for a particular soil requires further parameters such as moisture content or void ratio as well as OCR and the applied stress level. An alternative is to derive locally applicable values of am from back analysis of settlements. The lack of a theoretical basis does not invalidate the semi empirical approach and the situation may be considered as analogous to the measurement of undrained shear strengths by cone penetration testing. Initially values of the cone factor, Nkt' were obtained by comparative testing with laboratory or in situ vane tests, but now N kt can be derived analytically, (Houlsby and Teh). As yet the analytical values do not agree closely with long established empirical values and the process of reconciling the differences will continue until agreement is reached. 59 B4.4 Consolidation Characteristics from Cone Penetration Testing The problem of the prediction of embankment settlement is twofold in that the amount of settlement and the time it takes to settle are interrelated and of vital interest. If the settlement occurs rapidly, ie during construction, or at the other extreme say 50 years, then the amount of settlement is relatively unimportant. It is in the nature of things, however, that road, or rail, embankments on recent alluvial deposits fall between these two extremes so that reliable predictions of both amount and time of settlement are required. The methods of prediction of settlement rate are relatively well established and depend on consolidation theory and appropriate laboratory testing on undisturbed samples to measure consolidation parameters. The consolidation theory usually applied is that of Terzaghi (1943). There are a number of limitations to this one dimensional theory, viz it assumes: only one dimensional drainage; D'Arcy's law applies for any hydraulic gradient; homogeneous fully saturated soil; soil grains and pore fluid are incompressible; constant compressibility and permeability; linear and time independent relationship between effective stress and void ratio (strain); infinitesimal strain rate and flow velocities, and it ignores secondary compression. To overcome at least some of these limitations many adaptations have been developed, Terzaghi - Rendulic or Davis and Poulos (1972), or alternative theories eg Biot (1941). Essentially the former two are three dimensional, the second allowing different rates of pore pressure dissipation and settlement - and the Biot theory allows the uncoupling of the direct proportionality between dissipation of pore water pressure and effective stress. The use of the more refined theories, whilst philosophically more satisfying, do not seem to have produced any significant improvements in practice in the prediction of settlement rates. Significant over and under predictions are common and the potential errors are such that rational engineering decisions are extremely difficult on, for example, whether an embankment will require accelerated drainage, or even a structural solution, instead of construction in the available time. The reasons given for the discrepancies are many and varied. The more common are that sampling disturbs the sample sufficiently to cause major changes in the consolidation characteristics; samples are not representative of the real conditions; the 60 consolidation characteristics are stress dependent and the real stresses in the field are not modelled correctly; the theoretical consolidation model is not appropriate, and horizontal and vertical permeabilities are very different. In addition, it is recognized that the applied stress level relative to the preconsolidation pressure, 0Ve' is of vital importance and also that the accurate measurement of the latter is difficult in soft clays because of problems of sample disturbance. The rates of settlement in these smaller and larger strain zones each side of the preconsolidation pressure are different, hence it is essential to determine the coefficients of consolidation applicable for both zones. Despite these difficulties the consensus remains that consolidation time predictions can satisfactorily be performed through the use of coefficients of consolidation which would usually be measured by laboratory tests. A major purpose of the author's work has been to evaluate coefficients of consolidation from cone penetration testing. From the historical review given in B3 it can be seen that relatively few authors have addressed the problem of rates of dissipation of excess pore pressure around a piezocone in order to estimate coefficients of consolidation. Essentially they are Torstensson and Wissa et al in the 1975 ASCE Speciality Conference on In situ Measurement of Soil Properties, Raleigh; the 1980 research report publications by Baligh and Levadoux, and Baligh, Assouz and Martin, and the 1981 10th Conference ISSMFEj- by Franklin and Cooper, by Parez and Bachelier and by Jones and van Zyl. Franklin and Cooper referred to Torstensson, 1975, for interpretation; Parez and Bachelier put forward a cylindrical solution acting as a reverse vertical sand drain and having a radius R equal 4r where r is the cone radius. The drain theory gave a time factor T r of 0,03 for 50% dissipation and utilizing the expression between the coefficient of consolidation (radial) Cyr and the time for 50% dissipation, tf' as follows: B4.26 Parez and Bachelier (1981) indicate that for one of the two sites tested the agreement in cv for the piezocone, laboratory and backfigured from a site record were very close viz 1.0, 1.5 and 1.0 x 10-2 cm2/sec; whereas for the second site the piezocone and field results were 1,7 and 4 x 10-2 cm2/ sec respectively. The pore pressure was measured 61 along a cylindrical section of the penetrometer some distance above the cone so the system was not a piezometer cone in the sense that cone and pore pressures were measured simultaneously at the cone. The author, with van Zyl, (1981) eschewed a theoretical modelling approach and adopted a semi empirical method of direct correlation of piezocone measured, t50, (time for half dissipation) with laboratory measured, cy tempered with experience based on embankment settlement observations in selecting the appropriate laboratory measured Cy values. This resulted in the equation : where t50 is in minutes and cy is in m2/year B4.27 The temptation of having an easy to remember equation overcame the scientific compulsion to have consistent units and since in practice one generally measures t50 in minutes and requires the coefficient of consolidation in m2/year for calculation of embankment settlement times, the form of equation is convenient. The author, (Jones, van Zyl and Rust, 1981) justified this approach by using an idea based on Blight (1968) who estimated allowable vane shear testing rates based on consolidation of a sphere. If the penetration of a cone is stopped and no memory of how it arrived at the stop position exists, then the pore pressure dissipation pattern will be that for a sphere with the apex of the cone at the centre, where the surface of the sphere is at hydrostatic pressure. But how the sphere arrived is vital, because the residual effects are controlled by finite dissipation times. After penetration therefore, the pore pressure is not only that of the final position, but also includes increments from the cone's previous positions which may be considered as a series of stops at different time increments: these successive discrete spheres at different positions form a cylinder behind the final sphere of equal radius to the sphere. Thus, during steady penetration the cylinder dominates the pore pressure response, but that after penetration the final position and pore pressure response is predominantly that due to the equivalent sphere. The measured response, however, can only be measured at one 62 position (without having more complex cones) and since this is at the shoulder of the cone a compromise results. Theoretical studies by Levadoux and Baligh (1980) suggest that this concept of a spherical/cylindrical model and the position of measurement of the pore pressure are reasonable. The difficulty remains however of ascribing an appropriate value for the radius of the sphere and cylinder. There are three approaches to this: determine it theoretically; determine it experimentally or determine it empirically. When the author's work was being carried out to assess Cv from cone dissipation tests, viz 1978, no satisfactory comprehensive theory existed. The only experiments envisaged that could have led to the definition of the effective radius of the cylinder/sphere were penetration tests into samples containing numerous piezometers. Not impossible, but certainly daunting, particularly if a range of material types was to be explored. The third option of empirical correlation with other measures of consolidation times was therefore selected. Equation B4.27 implies that the function R1'f' ie drainage radius and time factor can be represented by a constant with in this case of value 50. It was accepted that the effective radius was dependent on the soil, but since the method was intended for use in the alluvial deposits of South Africa, and particularly of the Natal coast along which the geological history is consistent, this material dependence was not considered to be a significant problem. A large amount of laboratory test data and field experience was already available on these materials and the range of parameters for the more problematic materials was not extensive. Typically coefficients of consolidation for the soft clays are in the range of about 1 - 10m2/year and values higher than this are in any case generally indicative that significant long term problems will not arise since a large proportion of the settlement will take place during construction. This range of C v results in measured piezocone 50% dissipation times of 5 - 50 minutes which is acceptable for an in situ test. The author's 1981 equation B4.27 has been used for the past 10 years in South Africa and the efficacy of the method is demonstrated in Part D, the application of piezometer cone testing in South Africa. 63 The research reports by Baligh et al published m 1980 generally became available during the following year, and represented a major advance in the application of theoretical methods to the dissipation of pore pressure around a cone. The work is very comprehensive and it is not possible to give any but the most general overview. They defined cone penetration as an axisymmetric two dimensional steady state problem which is essentially strain controlled. Baligh (1975) working originally from experiment and theory on the penetration of wedges developed a strain path method of analysis which modelled the measured strains with considerable veracity. This method was then used to model laboratory measured excess pore pressures and their dissipation around cones in Boston Blue Clay. They indicated that the soil immediately around a cone after penetration has stopped is being loaded as the pore pressures decrease, but further away is unloaded as pore pressures increase before subsequently decreasing; also that accurate measurements of the ambient and generated pore pressures are essential if the dissipation rate is to be reliably estimated. Errors between measured and estimated, horizontal coefficient of consolidation, ch, occur at both high and low levels of consolidation and D = 0,5 gives the most satisfactory fit. 0 20 - 40 '" >-Q. W 0 6'0 80 100 Peal. Slill clay. Medium boston blue clay. Soft bolton blut -clay. ~ LABORATORY C. (NC) } ........-t LABORIrrORY C. (SW) CRSC TESTS (GERMAINE, 1978) DAVIS AND POUlDS I ~DUNCAN (MIT 1975) (MIT,197!1) ~ / I /' - ------- - ---J..~--___...=r---- /1' / I t-O-i/: .... // 1 BROMWELL AND ~ / I LAMBE (l96B) ~ / I KH/~ ______ ~ -----t-O-in--------HH C. (LOADIIIG)~ 1 I 0,0001 0 , 001 0,01 COEFFICIENT OF CONSOLIDATION, c,..(ZllOc . Ch (PROBE) Figure B4.12 : Comparison of predicted and measured coefficients of consolidation in Boston Blue Clay (Baligh and Levadoux, 1980) Figure B4.12 taken from their report illustrates that their method of estimating ch (probe) gives coefficients of consolida tion which agreed with Cv obtained by backfiguring from a real unloading case (an excavation). The estimated cn were about twice to five 1,0 64 times as large as those obtained during laboratory unloading and about twenty to forty times larger than the loading laboratory cy and than the two sets of site cy (loading) from back analysis of embankment records. Figure B4.12 shows good agreement between the carefully conducted laboratory tests to obtain cy and the backfigured site Cy. a correspondence that all too often does not appear to apply. Baligh and Levadoux (1980) show that from the general expression B4.28 and from virgin compression (normally consolidated) mv (NC) = -- log 1 + __ v CR ( !1a 1 !1 a v ave B4.29 and from recompression (over consolidated) mv COC) = - log 1 + __ v RR ( !1a 1 !1 a v ave B4.30 Using the conventional notation and where CR and RR are the compression ratios for the normally and overconsolidated ranges and hence for small increments of effective stress the above become ; ll\. (NC) CR = B4.31 2.3 ave ll\. COC) RR = B4.32 2.3 ave then RR ch (NC) = - ch COC) B4.33 CR 65 Assuming early consolidation around a cone is in a recompression mode then equation B4.33 applies: Ch (probe) = ~ (NC) and a yO = aye therefore ch (NC) = RR (probe) C (probe) CR h B4.34 or cy (NC) = RR (probe) ~ ~ (probe) B4.35 CR ~ The above equations result in a method of determining the required coefficient of consolidation C V (NC) for embankment loading directly from cone dissipation tests, but it is first necessary to determine : i) RR (probe) ii) ky/kh iii) ch (probe) from the field dissipation tests. i) RR (probe) : no theory existed (1980) for obtaining RR (probe) hence it can only be obtained for specific sites on the basis of measurements of the other parameters. From the tests available Baligh and Levadoux estimated that RR (probe) was in the range 0,5 x 102 to 2 x 102. RR (probe) is analogous to RR measured in a consolidometer, not having the same values but varying in the same range, which is fairly limited. Equation B4.35 is not therefore highly sensitive to the variation in RR (probe), and if the method were to be utilized in practice it would be anticipated that appropriate RR (probe) values could be readily estimated. ii) ky/kh has to be measured by appropriate tests. For any significant investigation of embankment settlement this will in any case be necessary to evaluate potential 66 two dimensional consolidation and possibly to judge the possibility of the use of sand drains. The range in the ratio for normally consolidated clays is fairly limited, say 1 : 1 to 1 : 2, and an estimate will not result in a significant error in assessment of settlement times for embankments. 1,0 0) ISO PROBE 0,8 " " o z '" :: ~ •• n= s;:: RANGE Of ch OUTfLOW ~ TESTS AND 110' SPEcn,IENS ~~:N~~ 100 1000 o 10 100 1000 VERTICAL EffECTIVE STRESS. CT; IkPal Figure B4.16 : Values of <1. from laboratory tests (Sills et al, 1988) In 1988, at ISOPT I, Sills et al described the consolidation of an embankment in Rio de Janiero and state that they used the Baligh and Levadoux (1980) method and compared the results of their piezocone data with laboratory and field consolidation data, the latter backfigured from field settlement measurements. Their data illustrates all too well the designer's problem, Figure B4.16 showing the spread of laboratory coefficients of consolidation from 3 to 300 X 10-4 cm2/s. 72 There is very poor agreement between their piezocone derived ch, the laboratory values and the backfigured coefficients, which may in part be because they do not take account of Baligh and Levadoux differentiation of stress levels for the estimation of c ie overconsolidated and normally consolidated ranges, and the v data indicates that the embankment stressed the subsoil through both ranges. Using only the recompression approach they obtain ch (cone) values of 90 -2S0 x 10-4 cm2jsec with an average value of 133 x 10-4 cm2jsec. The measured time for 50% dissipation is about 10 minutes so the Jones and van Zyl (1981) method gives a ~ of about 5 x 10-4 cm2jsec. This compares with the low end of the laboratory range given in Figure B4.16 (for an effective stress of about 2S kPa) and the Sills et al 133 x 10-4 cm2 j sec with the higher end of the range. The backfigured field measurements gave Cv values of about 12 x 10-4 cm2 jsec. If the normally consolidated Baligh and Levadoux corrections had been used, then values of Cv would have been very , much lower, probably about 3 - S x 10-4 cm2 jsec. This suggests that the emphasis given by Baligh and Levadoux on the distinction between consolidation rates in the normally and overconsolidated ranges would, if applied in this case, have given much closer agreement between measured and predicted coefficients of consolidation. In ISOPT I, 1988, Houlsby and Teh introduced another approach to the analysis of dissipation tests with piezocones. They estimated generated pore pressures using Henkel, (1959) : ~u = ~(Joct + a ~'toct B4.40 Using uncoupled Terzaghi-Rendulic consolidation theory solved by a finite difference method, they showed that the shape of the pore pressure dissipation curves is influenced by the rigidity index, Ir, since the initial pore pressure distribution is determined by Ir : this is because the excess pressures develop primarily in the plastically deforming zone the radius of which is a function of v'I r . They then showed that the dissipation curves can be unified in the range Ir of SO to SOO if a modified factor T· is defined in terms of Ir : 73 T* = B4.41 Figures B4.17 a) and b) are taken from their paper as is Table B4.3 which gives T· for the piezometer at a number of different positions on the penetrometer. Table B43 : Modified time factors T· from consolidation analysis Degree of Consolidation Tip Cone Face 20% 0.001 0.014 30% 0.006 0.032 40% 0.027 0.063 50% 0.069 0.118 60% 0.154 0.226 70% 0.345 0.463 80% 0.829 1.04 I ,0-'-=----=:::---- 0,8 Location Shoulder 0.038 0.078 0.142 0.245 0.439 0.804 1.60 1,0 0,8 5 rad. above shoulder 0.294 0.503 0.756 1.11 1.65 2.43 4.10 Ir valu8S : 50 to 500 10 rad. above shoulder 0.378 0.662 0.995 1.46 2.14 3.24 5.24 ;;- 0,6 B4.5.2 Piezometer cone soils identification The development of piezocones since the late 1970's created a new dimension for soils identification. Almost immediately the potential of the piezocone for identification was recognized. In 1980 Baligh, Vivatrat and Ladd discussed soils identification using the results of a standard cone and of a Wissa type probe (piezometer only) at adjacent positions. They demonstrated the potential of using a pore pressure ratio ut/qc (where ut is the developed total pore pressure and qc the cone pressure at the same depth). Since they were working primarily in one deposit, Boston Blue Clay, their interpretation was aimed primarily at assessing the overconsolidation ratio. They observed that the pore pressures measured during penetration followed a similar pattern to the cone pressures, except that in the layered soil system the cone pressures showed more distinct jumps between the layers than the pore pressures. They argued that in lightly overconsolidated clays undrained shearing results in decreased effective stresses which implies increased pore pressures not only to resist the penetration compressive stresses but also the large shear stresses. Conversely in heavily overconsolidated clays either smaller or even negative pore pressures will be developed due to shear. The compressive stress pore pressures mayor may not compensate for the negative shear stress induced pore pressure and will be related to the degree of overconsolidation. They summarized by stating that "The ratio ut/qc should provide a new promising method for soil identification. However, more data are needed to establish general correlations". In short, therefore, because of the limitations of their equipment and the fact that their work was confined to Boston Blue Clay and to Atchafalaya Clay, neither 77 having much variability in soil type, their observations although valuable, particularly with regard to overconsolidation ratios, were restricted. In 1981 (Stockholm and St Louis) the author published two papers which inter alia discussed soils identification. It was suggested that the Baligh, Vivatrat and Ladd (1980) usage of uJqc where U t is the total pore pressure is unsatisfactory and that the excess pore pressure, ue, should be used where : B4.44 and U o is the ambient pore pressure which would usually (no flow situation) be the hydrostatic pressure. The author suggested the use of a normalized subsoil index : U t - U o / qc - avo B4.4S U o avo The author noted that U e could be negative, and hence the index negative, indicating dilatant materials, and also that the index was Ita measure of the pore pressure parameter at failure, At". These papers coined the use of the description CUPT which subsequently succumbed to the arguably more rational acronym, CPTU. In 1982 (ESOPT II) and 1983 (Paris) the author presented his Soils Identification Chart -Figure B4.20 based on Ue and (qc - oYo) ie these were not normalized. The move away from normalized parameters was made with some reluctance, but there are advantages, viz, close to the water table and to the ground surface the normalized pore pressure and cone pressure parameters may become very large and cannot readily be contained within an arithmetic scale; it is useful to use the actual pore pressures rather than a normalized parameter since this results in developing a feel for the values; the chart is simpler to use. The normalized and unnormalized versions are in any case similar, since the normalising parameters U o and 0 YO are directly related and of similar values. ,0 , t kPc 400 200 100 o - 100 - 200 78 [ZJ Cloy .. lE] Clayey Slit. [[I) Sandy Silt. [ESand. Figure B4.20 : Soils Identification Chart (Jones and Rust, 1982) filter AZ----\ Lood Cell Pore Pressure Transducer }----AZ AI CROSS SECTION AREA A2 CROSS SECTION AREA CONE NET AREA RATIO • AYA2 Figure B4.21 : Piezocone net area correction The cone pressure, qc' should be corrected using the unequal end area correction suggested by Campanella et al (1982) and illustrated in Figure B4.21 and given in the following : where qT = qc + U (1 - a) qT = corrected cone pressure qc = measured cone pressure u = measured pore pressure a = net area ratio (see Figure B4.21) B4.46 The boundaries between the soil types on the soils identification chart, Figure B4. 20, are slightly curved since this is what the data appeared to show. However the boundaries should be seen as transition zones and these could be represented as 79 straight lines of constant values of : where B is defined as the CPTU pore pressure parameter. q The boundaries for the different soil types are given in Table B4.5. Table B45 : Bq values for different soil types Soil Type Bq Sand 0-0,01 Silty sand 0,01 - 0,03 Silt 0,03 - 0,09 Clay 0,09 - 0,5 B4.47 In practice it is found that at the low cone pressures found in soft and very soft clays (qc - avo) is small and the chart becomes difficult to use. A further version (unpublished) is given in Figure B4.22 with the cone pressure axis at a logarithmic scale, hence finer discrimination is possible at low cone pressures, and the soil type boundaries are represented by lines of constant Bq. 0 Il. ~ -a ::;, -' ::;, 0,8 0,7 0,6 O,!! 0,4 0,3 0,2 0,1 O,!! 2 3 4 !! ( ~c - (J'vo) M Po Figure B4.22 : Soils Identification Chart (Jones, 1992) 10 20 30 80 It is useful to examine the significance of the cone pore pressure parameter, Bq, not only as a convenient means of defining the boundaries between soil types. Peignaud (1979) using Vesic (1972) cavity expansion theory deduced that: where Af = Skempton's (1954) pore pressure parameter at failure. Hence at failure in a normally consolidated clay since Ar = 0,95 then: Then for typical values of Ir from say 50 to 500 : and from Au = (5 to 7.3) I Nkt 0" YO 0-' vc 1m I n. l ~=-r::~~J::",:::-~~O:Jr::' c:::o::=-r 0" vm Ca.CI4randel RECOMPRESSION INDEX , Cr SWELLING CURVE MINIMUM RADIUS SWELLING r INDEX , C. ef _____________ _ COMPRESSION INDEX Cc 0'" yl E..-FECTI VE PRESSURE. O"~ (LOG SCALEl Figure BS.2 : Terminology used for oedometer tests B5.4 Figure BS.2 indicates the conventional definition of the terms in the above expression. A second refinement has been the introduction of a method of calculating secondary settlements, ps. The most common method is given in the following equation : B5.5 where ell is the coefficient of secondary compression viz the change in void ratio per unit change in logarithm of time after the end of primary consolidation, tp. In the above expression ts is the time to which the secondary compression is calculated. 90 C Il would usually be obtained from consolidation testing although with the usual 24 hour load cycles this may not be possible and much longer laboratory testing times may be necessary. Alternatively CIl may be obtained from local experience or from published information. For example Mesri and Godlewski (1977) showed that there is a unique relationship between CIl and Cc for any soil and that CIl/Cc lies in the range 0,025 to 0,10; the higher values occur in organic soils. Davis and Poulos (1968) developed methods of settlement prediction based on elastic analyses in which moduli values, E, and Poisson's ratios, v, are required for the undrained and drained states to model the equivalent of the immediate and total settlements given by the Skempton - Bjerrum method (equation B5.3). The elastic settlement equations are : BS.6 1 Pt = ~ - (aa - v'(aa + aa ») az L.J E' Z x y BS.7 An important condition of the above is that elastic soils are defined as those in which the settlement is independent of the stress path viz the total settlement is not dependent on rate of loading and will be the same whether the embankment is built in many stages or one stage (provided no overstressing occurs with the latter). In other words the final components of immediate and consolidation remain the same whatever stress paths occur. It is emphasized that equation B5.7 represents the total settlement and therefore includes the component due to the immediate settlement given by equation B5.6, ie equation B5.7 is equivalent to equation B5.2 the Terzaghi settlement derived from consolidometer test results. For homogeneous elastic soils the stress-strain drained behaviour is defined by Young's Moduli, E', and Poisson's ratio, v' and a consolidometer test gives: BS.8 91 Note that Il\r and liE' are not equal even for the relatively simple case of elastic homogeneous soils and this distinction is important and often overlooked when the derivation of compressibility parameters from in situ tests is discussed. For homogeneous elastic soils the undrained and drained elastic moduli can be related through the shear modulus G viz 2G = E' B5.9 = 1 + v' and for undrained clays Poisson's Ratio is 0,5 therefore: E = 3 u E' B5.1O 2(1 + v') Davis and Poulos (1968) estimated the relative value of immediate undrained settlement to total settlement for a range of drained Poisson's ratios and geometry - Figure B5.4. They also estimated the error involved in using the conventional one dimensional approach instead of a three dimensional analysis - Figure B5.3. From Figure B5.4 for road embankments where hi a is about 0,5 - for a typical single carriageway main road over alluvial deposits on the Natal coast - then for soft clays if u' is 0,4, the proportion of immediate undrained settlement may be about 0,35 of the total. It is also noteworthy that this proportion is highly dependent on the ratio of embankment width to subsoil depth. Similarly, from Figure B5.3 for the same embankment, the calculated one dimensional settlement should be increased by about 15% to derive the equivalent three dimensional settlement. I,O~;::--_=---_--==~!:....:..~ i ; o Q .. ~ ~ 0,. o :z: ~ .. .. ~ z .. ... :z: :II .. ~ ~ 0,' ~ ~ OJ .. ., OJ oJ ., : ~ 0,4 ~ ~ z .. .. oJ > '" g ~ 0,2 o H I + O~---'---r---r--~ o % Figure B53 : Error in settlement for one dimensional approach (Davis and Poulos, 1968) 0 1,0 v'. 0,5 .. :i .. :l z .. ... oJ," ~ ~ OJ .. ., .. w - ., 0 Of Z oJ - '" "'z 0: _ 0 .. z :> - oJ t! ~ '" 0 - .. e 0, :II ;! 'I. Figure B5.4 : Relative importance of immediate settlement (Davis and Poulos, 1968) 92 Measurement of Eu is difficult and therefore values must either be obtained from equation BS.10 which requires a knowledge of the value of Poisson's ratio, or values of E can be estimated from E Ic relationships such as shown in Figure BS.S. It would u u u be useful to have values of Eul Cu from back analysis of settlements for local soils; unfortunately such data is not readily obtained other than from detailed research level investigations.r--_________________ • " o "- " ... 1000 Z 348678910 OVERCONSOLIDATION RATIO Figure BS.S : Ratio of Eu/cu against OCR for clays For anisotropic soils the equations BS.9 and BS.lO should be modified to account for differences in the horizontal and vertical Poisson's ratios and the resulting estimated settlements may change by 20% for high ratios of E/h/E/ v ' Burland, Broms and de Mello (1977) discuss these points extensively in their state of the art of settlement predictions for foundations, but not for embankments where the factors of safety or stress ratios will generally be much higher than for structural foundations. They conclude that because of all the complications, both theoretical and practical, the elastic methods have no advantage over the Terzaghi approach in giving accurate predictions of settlement. For embankments on soft clays there is a further complication in settlement prediction which is the result of the relatively high stresses imposed compared with the in situ stress. Local yield occurs which is non elastic and cannot be predicted on the basis of 93 elastic type parameters. In order to predict deformations it is necessary to use suitable (stress-strain) relationships and numerical methods. o ..,....--......... -""""'"-00::---..:::----, ~ l.o-r---c---.....:::--......;:'~"'IIIiO:----, a: Q;-o,8 ... , Il- 0.6 ~ ~ ,-~,6 ~- ... ~ (c) '.0.4 ~II 0 II 0.2 SOIL LAYER (0) till) 0 II/B a l." In O+--~-~-.--.--~ In +--~-'--~--r-~ o 0.2 0.4 0.6 0.8 1.0 0.2 0.4 0.6 0.8 1.0 o APPL.IED STM:SS RATIO. q IqUIt APPLIED STRESS RATIO q/qult 5 a: 1.0 -.--...---:--.......::-~..,----, lE 0.8 ~ 0,6 1&1 ~ 0.4 ti 0.2 II) (bl 11/8= 1,0 O+--,--,--.--.--~ o 0.2 0.4 0.8 0.8 1.0 APPLIED STRESS R,u'IO. qlqult Figure B5.6 : Relationship between settlement ratio and applied stress ratio for strip foundation on homogeneous isotropic elastic layer (D'Appolonia et at, 1971) D'Appolonia et al (1971) developed a method to obtain a settlement ratio SR defined as Pie/ Pi: where Pie is the immediate elastic settlement and Pi is the actual immediate settlement including local yield. SR is obtained from sets of curves -Figure BS.6 relating SR' the applied stress ratio q/qult and f, the initial shear stress ratio. The initial stress ratio is defined below and can be obtained from Figure BS.7 : BS.11 I - a/ho a YO f = 2 Su BS.12 1 -K or f 0 = 2 SU/a'yO For soft normally consolidated clays f is usually in the range 0,6 to 0,7S, and for embankments along the Natal coast h/B is about D,S. Such embankments usually have a stability problem so q/qult is at least 0,6 and may well be up to 0,8. 94 From Figure B5.6 it can be seen that for these conditions, and with f of say 0,65, then SR is 0,6. It is therefore clear that local yield may have a significant influence on the total settlement. ---0 _ 0 0 ~ '-I :0 ~ <..l N 0.8 0 .6 04 0.2 0 -0.2 SOIL PI (%) SOURCE Boston Blue Cloy 21 MIT Skempton Weald Cloy 24 Sowo (II Varved Cloy 130(bulk) j MIT IMe Organic Cloy ! 3B MIT j Domes , AGS CH Cloy is Moore[2' Bangkok Cloy i 41 IAIT S MIT 1 [2,197 5 I I 111 1963 / La~ PI j j I Note: All data for 'I unloading f ~om o;m 2 4 I I OCR = ~m lO'vc I 1\ I 6 8 10 Note l cu = 0 .5 (0; - 03), from CK.U triaxial or Dione strain tests . K. from Brooker and Ireland (1965) for Me. Organic Cloy. I ,0 -I:-L-..----1 __ --L __ --L-__ ..L.-__ L---1. __ -t 0,_ : Ote 2 .. ':. 0,. .. 0,2 0+-.--.---,---.--.---.--.---+ 0,25 0,3 0,. o,tS 0,6 O~1 o,e o~ 1,0 SHEAR STRESS I SHEAR STRENGTH froin piezocone dissipation tests by correlations with observed rates of embankment settlement. Part C describes the first and Part D the latter two. 97 PART C : SOUfH AFRICAN DEVELOPMENTS IN CONE PENETRATION TESTING C1 INTRODUCTION An historical review of international developments in cone penetration testing was given in Part B so that a background could be provided against which South African developments could be assessed. Section B2 described the mechanical cone systems from those in the 1930's to the present day. Section B3 covered the development of electrical piezometer cone systems from their beginning in the late 1970's to the present and included mention of the author's contribution to this. Section B4 discussed the interpretation of cone penetration testing with the emphasis on parameters required for the design of embankments on soft alluvium where large settlements and long consolidation times give rise to significant engineering problems. The author's contribution to interpretation methods is again mentioned. Part B5 described the application of cone penetration testing for the design of embankments and is a review of current international practice in settlement and consolidation time analyses. Part C has a similar format to Part B of an historical reView, C2; followed by mechanical cone penetration testing, C3; through a description of a consolidometer cone apparatus developed by the author, C4; to piezometer cone developed by the author, C5. The application of cone penetration testing by the author is described in Part D. C2 SOUTH AFRICAN MECHANICAL CONE PENEfRATION TESTING (1950 - 1975) There have been few publications by South Africans on cone penetration testing and hence it is relatively simple to trace the history of development of the method locally since the author has been directly involved since 1965. This does not suggest that the system has had little application locally, since the merits of cone penetration testing 98 were appreciated in the early 1950's by Kantey (1951) who advocated its use and was instrumental in introducing the system to South Africa. In the early years the local system was essentially very similar to that in use in Europe with the exception that smaller diameter equipment, casings and cones, were used. This was because the drilling industry manufactured as standard, E size drilling rods of 33 mm diameter and cones were made to this size, hence the cross section area was smaller than the standard European 1000 mm2. The rigs used in South Africa were purpose made locally and similar in concept but not in detail to European machines. The local machines generally used hydraulic rams and pumps from earth moving equipment. No automatically controlled penetration rate was possible and the rate was determined by the judgement of the operator. Although not sophisticated this was adequate for the purpose, which was primarily for pile design. However a less desirable feature which persisted for a number of years was that the load required to push the cone and rods was usually determined by measuring the hydraulic pressure in the main operating ram and not by a separate load measuring system. The accuracy was very poor, particularly at low loads, and uncalibrated pressure gauges exacerbated the problems. The primary purpose of cone penetration testing at the time was as an economic and relatively accurate alternative to boreholes with Standard Penetration Tests (SPT) for determining the depth and density of sandy subsoils for piles. Such strata would probably require SPT-N values of a minimum of say N = 15, for piling, which is equivalent to a cone pressure qc of about 7 MPa. A 100 kN CPT rig would generally be limited to a maximum of say 40 kN load on the cone, ie 40 MPa, but the pressure gauge would have to measure the full load pressure on cone and rods of 100 kN which , , with a suitable safety margin, leads to a load measuring requirement of 150 kN. The hydraulic rams were usually about 100 mm diameter and hence maximum gauge pressures of about 20 MPa (3000 Ibsjin2) were common. For firm to stiff clays, or loose sands where cone pressures are less than 2 MPa, giving gauge pressures of less than 0,25 MPa, ie about 1 %, or 1 division, of the full gauge reading, the accuracy of the system was poor. For soft clays, where cone pressures of one tenth of these could be expected, the system was totally inadequate and in soft to firm clays zero readings were often recorded. Figures C2.1 and C2.2 are records from 1953 showing calibration of f~O :. ((too . (DQ . . ~~ .• . : ~~ , : ~~"It' tl/ .IJft".J \-., ~~~~~ .' 'r . .ll" JF 1~ ~r : DI "'JI~'I J N.t:! i' .~ . ~ ,. . -fi- .::c '. ~JC,~~-~ ~~" 1\,) ~';;;~;;~ ~~'~? ' . -:. );.;\,~,~_:-: to/I? .~~.~~¥~:t:!::~~.~~.~ /)"TCH l'£olJc(S CO/,/YE&S/ON 7RdL.e,5 I . O I Rf. R~f/"/Nq 7"0 f. as/o,? "'0$. - _. ! ~ ;;", Po/.-./, . tf'et~/.s~Il~Ce . 1-- . , ~Pg \!>ooo 12Soo I_ ~ 1_77S 11'130 1 3<1- 2<130 , 7300 ! t,-9 b-oCo I/Z'~oo' 10 , ~ bo i 215-0 I J;; 3010 : 7$"10 I 60 1 :i170 1~900 L~- ;;: :~::: I ;; :::l ::;: \ ~~ \:~:: I~~: tl'-I(;~-I' -I(::Oo i 3 0 00 J"I :33bO gjbO \ 6'1- \ .-s-~o \IJ700 , • <-7' i 3ZZo 40 3440 "boo t &' ,S60o 11700 :. '~{, \' /370 13430 4 1 ,nO 'Iioo i {'6 ' S6to 14100 ' 17 : /460 1 36So I oSOO I :~ " 0 <0 ""'NT RESISTANCE '" 20 '" <0 0 I \ 2 \ '~ .... '11 . \ . 4[> \ • • i 7 r;- - I, 8 1\ -I', · ...... , 10 R Ls " \ 12 13 " I " " 17 18 19 1 20 1 IIkNI TOTAL RESISTAJ.IC( ( ... ~ I ", ' 60 70 10 20 >0 <0 ", ' 60 70 10 (11."'/",2, 10] 1 I II.N/",2. (0) 1 ""'NT RESISTANCE '" '" 60 70 '" 20 '" <0 '" 60 70 0 I 0 , <:. Peonl <. PoInl , I R.,.,tonel I RUI,IO",t '-'"I : 10tol 2 " . Toiol 2 .: -'. < .. .. RUI,lone, .. Ruuloncl , , , -" -4~ 4 "- I , ,.- I , , , "£ - "£ • , - • - .J 7 .J 7 ~ ~ i:. ~ .J 8 B , Q ~ . --- .............. 5 • :i1 0 <" " ;:; ., 10 10 C . ~ )) g " " f) <0 " I 12 . 12t--- 'J S \ E i 13 13 0 0 / .. .. J .,1\, -- <. -- " I.I( \ I. \ : 17 / 1 17 ( '\ \.... , 08 j o. " -19 o. 1\ I 20 10 TYPICAL LAYERED SOFT TO FIRM CLAY AND LOOSE TO DENSE SANDS MEDIUM TOTAL RESISTANCE , kNI 20 >0 <0 '" 60 70 PONT RESISTANCE ( II.N/",2 , ",' 20 >0 <0 '" 60 70 I <. PD.n. R.",tonel Totol -',j < ' .. R,IIIIone. G } 1 1 . I - I .. , \ I , ..- , - c, I I '- t ~ I \ ,. < - , - - --- Fig. C· 2· 3 ( b) SAND hne white with cloyey loyers SANO f in~ clQy~y whi l~ SANO 'in~ grey SANO me-dium r in~ ol .... ~ SANO medium ctoy~y yellow C LAY WIth shells g~y SANO med .... m ctay~y gr~y SANDSTONE B cretoce-ous .. , ., 5 Il T fine sondy brO'w'n SAND medium grey SA.ND mt'dium cloyey grey SAND fjne silly block Cl AY silty black SAN 0 I jne grey SAND silly grey SAND c1Qy~ ~Iow SANDSTONE soft wtot~r~ hard Inxhxld hard intoct D Freeway Inlerchange sile 01 Dalbridge Durban Tonk Inslallalion siles near Airport Durban TYPICAL LOOSE TO MEDIUM DENSE SILTY AND CLAYEY SAND Fig. C·2·3 (0) 102 differential settlements of the wall units was an important factor in the design. Due to financial restrictions the investigation for the project took from 1961 to 1965 and in this period thirteen boreholes and fifteen CPT's were put down as well as other sampling holes and inspection shafts. It was recognized that as valuable as SPT's and CPT's were to provide in situ assessements of compressibility, it was necessary to calibrate these against other tests. Although undisturbed sampling of the more clayey soils, followed by laboratory testing was feasible, most of the subsoil comprised cohesionless soils which precluded this approach. The City Engineer's Department and Kantey and Templer, assisted by the Building Research Station of the Council for Scientific and Industrial Research, conducted a very large plate loading test - Figure C.2.4 - and small diameter screw plate tests. The description of the tests given in the following two paragraphs is condensed from the project report in which the original photographs appeared. The plate was a reinforced concrete 6,1 m square slab which was loaded with 800 tons of pig iron giving a pressure of approximately 200 kPa. CPT's were put down before and after the test at positions immediately surrounding the plate and through holes left close to the centre of the slab. Precise levelling of the corners and centre of the plate was carried out during the loading, for 5 weeks after loading, during the subsequent unloading and for 6 weeks after the removal of the load. It was intended to leave the load in place for a longer period to give a better measure of longer term settlement but the pig iron load was required for export. Nevertheless, the rate of settlement data indicated that the settlement after full load was only about 4 mm in 5 weeks compared with 75 mm during the loading stage - Figure C2.5. The screw plates were 6, 9 and 15 inch diameter; the tests were at depths of up to 18 m (60 feet) and loaded to pressures of 1,7 MPa. Settlements varying from 2 mm to 70 mm were measured at various stages of loading. At the time the method was generally perceived to be very promising in that it allowed a practicable means, in materials that were difficult to sample, of measuring in situ compressibility which fundamentally was more satisfactory than cone penetration testing. However, reasonably practicable though it may have been, screw plate testing could not compete with cone penetration testing for convenience, economy and simplicity of operation and had, until a recent revival, (eg Bergado et aI, 1991), more or less vanished from the scene of site investigation techniques. ~~tfIT.fF~;,~~ \'. " , .. f - DALBRIDGE PLATE LOADING TEST (1965) Fig. C·2·4 UNIFORM LOAO-lhs.persq.ll. 000 n===~ __ ~m~OO~ __ ,-__ ~2~OOOL-__ ~ __ 23TOOO~ __ ~ __ ~~~ 09... --- -- .::::- i-----=-='_==-I--------~- Norlhcornu -.-- .. -.-- -- :::--=-= ~--+----k---~ Wlc~ -- --- -- ---- ---- Wt's/ corner Soul h corner J I-----j-----t- --.. - ----- - -I----j--"\-t- ---I -I------t-----!--~cent'e LOAD SETTLEMENT CURVES FOR 20 FT. SQUARE CONCRETE SLAB Fig. C· 2 . 5 (a) TIME - Days (Aft", reaching f,,1/ /OtJd) ~ __ -T-5 ____ ,lrO ____ i5 20 75 .10 .15 ., · 051------'~;:__-~~:__I------t--_t_---_t_--_____j '" -<: o ~ h.. ~ ·/01-----1- --1 ~ ~ I , "-< .,., ./51------+-----j----/-----t----j----="'t""':::::::----j ·20L--__ --'-____ -'---__ -----'-____ --' ___ --' __ __ ~ __ ~ PROGRESSIVE SETTLEMENT OF CORNERS OF A 20 FT. SQUARE CONCRETE TEST SLAB UNDER A UNIFORM FULL LOAD OF 4,000 LBS PER SQUARE FOOT Fig. C·2·5 (b) lOS The Dalbridge Flyover project, and particularly the very large plate loading test, enabled correlations to be made between measured settlements and predicted settlements from both laboratory and field testing. In 1961 Schultze and Menzenbach published correlations between compressibility and Standard Penetration Test N values derived from sampling and laboratory testing. These correlations were expressed in regression equations in the form:- where and 11m, = 71 + 4,9 N N is in blows 1ft m, is in cm2/kg C2.1 The above was for fine saturated sands and similar expressions were quoted for other materials. This form of equation was used by the author for the Dalbridge results and modified to generate a local correlation for clayey sands : 11m, = 18 + 4,4 N C2.2 In the Dalbridge report SPT N values and CPT qc values were directly correlated, viz qc (tons/ff) = 3N, but no equivalent equations to the N value ones given above were stated. Cone pressures were correlated with compressibility using the then conventional · relationship 11m, C2.3 The SPT equation gave settlement predictions which fitted the measured settlements better than those using the CPT data and equation C2.3 to derive compressibilities. The report stated that at this site predictions of settlement using the SPT values were more reliable than those using the CPT data. This general conclusion, however, was not justified since the SPT equation was fitted to the settlement data by adjusting the constants, whereas the CPT equation used only the unmodified qc and m, relationship in equation Cl.3. Kantey (1965) referred to this work at Montreal. Webb (1969), again described the Dalbridge Flyover work and that at other sites, and quoted two correlation equations for compressibility and cone values which were 106 derived by the author from the Dalbridge results viz: E(m2/kN) = 5/2 (qc + 3000) for fine to medium sand C2.4 E = 5/3 (qc + 1500) for clayey sands C2.5 It was noted that the correlation of N values and cone pressures at one of sites which was an oil storage tank farm, gave:- qc = 2,2 N (qc tons/ft2). C2.6 Webb concluded this paper by remarking that "more reliable results are obtained from the deep sounding test," a different emphasis from that given in the Dalbridge report. The change can be ascribed to there being both more direct local experience gained in the intervening time and to more international experience being available through the literature. Webb and Hall, (1969) described the use of cone penetration testing to monitor the efficacy of vibroflotation at a number of sites in the Durban area, viz the Durban Sugar Terminal Silo, a Factory Site and an Oil Tank Site. The author was intimately involved in this work and both advocated and controlled the cone penetration testing which determined the pattern of the vibroflotation. At the Sugar Silo Site a total of about 50 CPT's were carried out. Initially CPT's were at varying distances from trial vibroflots to assess the increase in density so that the required spacing of the virbroflots could be designed. The CPT's were then used as a control test on the vibroflotation during construction. Similar but less vibroflotation using the CPT as a design and construction control system was carried out at the Factory and Oil Tank Sites. For the Dalbridge site and for the Sugar Silo work the procedure suggested by de Beer was used for the intepretation of the CPT results, ie they were expressed in terms of the friction angle 121 which was calculated from the cone resistance and the assumed overburden pressure - see Figure C2.2. The friction angle was often shown on the logs and was seen as the definitive result of the CPT. Vibroflotation was then specified by requiring a minimum value of (I). However, the procedure for estimating (I) was recognized to be approximate and valid only for sands. Where clayey strata or lenses were encountered in the sands the cone resistance, and hence calculated 0, were much 107 lower so that specifying a minimum envelope for 0 was impractical since it would require different values for each material. The procedure was therefore changed to specifying a minimum cone resistance and accepting lower values in clayey layers. The work at these sites in the Durban area in the 1960's gave considerable impetus to a more general acceptance of cone penetration testing for subsoil investigation not only for pile design but for the estimation of settlements. In the period 1969 - 1973 the author was primarily involved in the geotechnical investigations being conducted for the development of the national road system in Natal which is described in Part A. Since it was apparent that highly significant problems would be encountered with embankments over the estuarine deposits, the author decided that improved geotechnical investigations would be necessary and that cone penetration testing could fulfil an important role. The conventional investigations then, as now, consisted of boreholes with undisturbed sampling for cohesive materials, followed by laboratory testing, and boreholes with Standard Penetration Testing. The latter, in the softer deposits gave N values in the range 0 to 5 and it was apparent that the results were very dependent on the operator and equipment, particularly that for raising and dropping the SPT sliding hammer. To overcome the operator and equipment problem for the SPT the author introduced the use of automatic trip hammers. The method previously in vogue was based on American practice and consisted of a rope wrapped around a winch on the drilling machine. As has often been recorded, Fletcher (1965), the method suffered the twin drawbacks that the hammer did not necessarily fall freely and that the height of the drop depended on the operator's skill and diligence. Automatic trip hammers worked on the principle of a mechanical latch which was released when it passed over a larger diameter section of the guide rod thus dropping the hammer. A disadvantage of these systems is that if the lifting cable is co axial with the drill rods, which is mechanically ideal, then head room above the anvil and hammer assembly may be very restricted unless a tall mast or tripod is used with the drilling rig. If, on the other hand, lifting non axially is accepted to reduce head room problems, then lateral loading on rods causes other problems. The author developed, and used successfully for a number of 108 years, an electro-magnetic trip hammer. Figure C2.6 shows the device which worked off the drill rig 12 volt battery. Different switch arrangements were used with various degrees of success : (a) in which the current was interrupted by contacts passing over an insulated section of the hammer guide, and (b), in which an industrial switch was operated by passing over a smaller diameter section of the hammer guide rod. The system worked well and undoubtedly played a part in changing the site investigation industry to using automatic trip hammers, although of the mechanical and not electromagnetic type. ~ SEE DETAIL FOR U . ./ RELEASE SWITCH :Y POWER CABLE EL ECTROMAGNET SPT HAMMER RELEASE SWITCHES (0) ELECTRICAL ( b) MECHANICAL NYLON BUSH SPRING LOADED SLIDING CONTACT Figure C2.6 : Electromagnetic SPT trip hammer In the course of developing mechanical and electro-magnetic trip hammers numerous tests were made to compare the automatic trip hammer with the rope over a cathead system. It was found that the blow counts for the two methods were significantly different, there generally being about 25% and sometimes 50% more blows for the manual system than for the automatic. However, despite the overall improvement in Standard Penetration Testing, the method was of little use in the more clayey estuarine deposits other than as an indication of which strata required more sophisticated investigation. 109 The author adapted a diamond drilling machine so that it could also be used for cone penetration testing. This entailed fitting an improved hydraulic flow control valve so that the penetration rate could be more finely controlled and an improved hydraulic pressure measurement system comprising a high and low pressure gauge. The drilling machine was also equipped with screw augers which could be drilled in to act as holding down anchors. A further improvement was the importation from Holland of standard 60° mantle cones (35.7 mm diameter) and the use of stronger EW (35 mm diameter) rods with a taper screw thread pattern instead of the lighter EX rods. Whilst these improvements accentuated the importance of standardised techniques there were still deficiencies. The penetration rate control was very much improved - and it was then generally thought that penetration rate was more critical than has been subsequently shown to be the case - but the system continued to rely on measuring penetration resistance by reading the hydraulic ram pressures albeit through a high quality double gauge system. The author therefore developed a method based on that used in Europe of a closed circuit hydraulic load cell. In essence this was simply a commercially available hydraulic jack connected to a twin gauge pressure measuring unit. Adaptors were made for the jack so that it could be fitted to the cross head of a drilling machine or to the ram of a penetration test rig. In the former case the adaptor was a socket end of a standard (N) sized drill rod which could then be screwed onto a rod and held in the drill chuck . . The jack and connected gauges were calibrated in the laboratory and the calibration checked from time to time. The system worked well and was assembled from readily available pieces of equipment. As a result it was specified for all investigations for the Natal Roads Department and has subsequently become normal practice in South Africa and at about the same time it was specified that the internationally recognized cone size (10 cm2) should be used. In 1974, Webb reporting on South African practice to the First European Symposium on Penetration Testing, ESOPT I stated that the older E-rod equipment was still in general use and the system of measuring pressures from the main hydraulic ram. This, 110 however, although a fair description at the time it was written, was already outdated in some respects by the time it was published with regard to the equipment being used. The comments on the usage of cone penetration testing are nevertheless valid and when these are compared with reports by other international contributors at ESOPT I it can be seen that South African practice, compared favourably with that anywhere else in the world other than in western Europe where it originated. Despite this however it was abundantly clear that the available cone penetration equipment both internationally and locally was inadequate for the investigation of the softer clays found in the estuaries along the Natal Coast because the load measuring systems were unable to measure the low cone pressures required. For example a subsoil with an undrained shear strength of 15 kPa would be expected to give a cone resistance of about 200 kPa. The usual dual gauge measuring units consisted of a 0- 100 MPa high load gauge and a 0 - 10 MPa low load gauge. A cone resistance of 200 kPa is therefore less than 1 % of the full scale low gauge reading - depending on the area of the load cell. Since a field operating gauge of this nature is unlikely to be more accurate than say 1 % of full scale reading, the accuracy of assessment of shear strength was hardly sufficient to allow any design decisions to be made other than that further investigation by some other means is essential. Nevertheless the inhomogeneity of the estuarine deposits meant that a relatively inexpensive near continuous testing method such as cone penetration testing was in many respects ideally suited for these subsoils. This, and the fact that the problems of road embankments on soft soils were recognized as a major difficulty for the design and construction of roads, led to the author being invited to the National Institute for Transport and Road Research in Pretoria early in 1974 to develop cone penetration testing. 111 C3 MECHANICAL CPT EQUIPMENT AND INTERPREfATlONDEVEWPMENTS IN SOUTH AFRICA During the period at NITRR (1973 - 1977) the author published a number of reports and papers incorporating the use of cone penetration testing and this section comprises a summary of these and of the research which provided the information. C3.1 Methods of Estimating Embankment Settlements using CPT The first of these was research report RS/6/74, Jones (1974) Methods of Estimation of Settlement of Fills over Alluvial Deposits from the results of Field Tests. It is a description of international cone penetration testing at that stage and a compendium of the methods of settlement estimation based on correlations of cone resistance with compressibility. These correlations are taken both from international and South African experience and include correlations of CPT qc values with the Standard Penetration Test, which was then much more familiar in South Africa. Essentially two approaches were adopted for the estimation of settlements. The de Beer and Martens (1957) method, and the Terzaghi based consolidation equation using the coefficient of compressibility, my; these are briefly described in the following subsections. C3.I.1 de Beer and Martens The subsoil is divided into an appropriate number of strata on the basis of material type or ranges of cone resistance. The average cone resistance for each layer is estimated from the CPT log; the overburden pressure at mid layer depth is calculated, usually from an assumption of subsoil densities, both above and below any water table, and the increase in pressure at the mid layer depth due to the imposed load (embankment) is calculated using a Boussinesq stress distribution method. s/H = l/C In (ayO + az)/ ayo) C3.1 where s = settlement of layer H = thickness of layer C = compression modulus 112 = overburden stress at mid layer Oz = embankment stress at mid layer The compression modulus, C, is given by : C = where = average cone pressure in layer. de Beer and Martens specifically referred to upper limits of settlement primarily so that decisions could be made regarding the need for piling of bridge abutment. This and subsequent experience on sands, for which the method was derived, led to Meyerhof (1965) and Schmertmann (1970) suggesting that it should be modified to C = C3.2 C3.1.2 Coefficient of compressibility, II\. This method uses a direct relationship between cone pressure, qe' and the coefficient of compressibility, II\. and the conventional Terzaghi compression equation: s/R = and = where = Oz II\. I/IIn8e constrained modulus coefficient C3.3 The constrained modulus coefficient 11m depends on the material type. In this method the material type would be defined either from boreholes and sampling or from friction ratios obtained from the cone penetration testing. The method is otherwise similar to that in C3.I.1 the subsoil is divided into layers and the settlements for each layer are calculated on the basis of the average cone pressure and embankment pressure within each layer and summed to give a total. The constrained modulus coefficient requires to be assessed for each layer. The publication RS/6/74 described the use of the friction sleeve in detail and the purpose of this was to encourage the use of the standard cone together with the friction 113 sleeve. The data given for Mtwalumi - south coast Natal - where the site work was conducted - reflects the first published use in South Africa of the friction sleeve and hence of friction ratios and materials identification by cone penetration testing. It had become practice in South Africa for CPT readings to be taken at 0,5 m or even 1,0 m depth intervals. Whilst this may well have been satisfactory when assessing sand densities for piling, it was inadequate in multilayered sands, silts and clays. The document therefore recommended that the depth interval should be not greater than 0,25 m and that this was necessary and convenient for assessment of friction ratio. This is so because the sleeve is approximately 250 mm above the cone so the calculation of friction ratio should take this depth difference into account. F.R. = Sleeve Pressure Icone pressure % However since the sleeve cannot operate independently but only in conjunction with the cone, the sleeve pressure is obtained by subtracting the cone gauge pressure Gcz from the cone plus sleeve gauge pressure Gcsz. This value is then related to the cone reading at the previous depth interval, ie 0,25 m higher taking account of the areas of the cone and sleeve (1000 mm2 and 15000 mm2) F.R. % = 6,7 (Gcsz - Gcz)/Gc(z _ 0,25) C3.4 Manual recording of gauge readings at 0,25 m depth intervals was tedious and the author believed that automatic systems could be used. It was common locally to record the pressure required to advance the string of rods both with and without the cone so that even without the friction sleeve each depth increment required three gauge readings. The report recommended that even with good CPT data settlement estimates should be expressed in such a way as to reflect the confidence in the accuracy of the estimate Ie Mtwalumi embankment settlement = 0,7 ± 0,2 m 114 No data existed in South Africa for the correlation of friction ratios with material type, hence it was recommended that a simplified version of Begemann's (1953) correlation for Europe should be used as given in Table C3.1. Table C3.1 : Material description from friction ratios FRICITON RATIO % MATERIAL DESCRIPTION 0-2 sand 2 - 2,5 silty sand 2,5 - 3,2 sandy silty clay 3,2 - 4,0 silty clay 4,0 > clay From these derived material descriptions, the cone pressures and published relationships (Bachelier and Parez, 1965; Gielly et aI, 1970) the constrained modulus coefficients, 11m are obtained. Alternatively the South African correlations of qc directly with Il\. can be used, equations C2.4 and C2.5. RS/6/74 also described the derivation of undrained shear strengths for clays using the conventional equation: = C3.5 and notes that at that stage no evaluations of Nk for South African clays were available but that the internationally generally accepted value of Nk = 15 for normally consolidated clays appeared to be satisfactory. Thus, for initial conservative assessment of stability, it was recommended that the undrained shear strength should be given by: = C3.6 The report noted that general relationships between coefficients of compressibility and undrained shear strength have been postulated for clays from normally consolidated to overconsolidated, (Skempton, 1951) 1/Il\. = (25 to 80) Cu normally consolidated C3.7 = (70 to 120) C u overconsolidated C3.8 115 and that if comparisons are made of derived 11\ values using the various approaches then a large range of values may be obtained from the same cone, qe' data. The remainder of document RS/6/74 describes in detail settlement from cone penetration test results for a particular site in order to demonstrate the method. A settlement estimation chart was devised by the author to allow the rapid estimation of settlements - Figure C3.I. This is based on the de Beer and Martens method with an cx of 1 5· other values of cx may be selected including those resulting from m " m equations C2.4 and C2.5. The chart assumes a fill density of 20 kN/m3 but if the density is different an adjusted fill height can be used. The settlement of layers can be individually estimated by subtracting the settlement to the top of the layer ie all material above it assumed to have the same properties, from the settlement to the base of the layer. A secondary purpose of the chart was to illustrate the marked dependence of settlement estimation on the values of cx m' hence the inappropriateness of detailed calculations. C3.2 Improvements to CPT Equipment - Vane Shear In view of the progress being made with cone penetration testing the NITRR purchased a CPT rig from Goudsche Machinefabriek B.V. of Gouda, Holland, who had been the principal developers of CPT equipment over the previous 40 years. The availability of the new equipment in 1974 enabled specific research projects to begin. The first aspect of this concerned the use of the friction sleeve and this led to the second aspect of improvements to the load sensing system. Both of these are reported in a paper by the author (Jones, 1975) to the Sixth Regional Conference on Soil Mechanics and Foundation Engineering held in Durban, September 1975. The first aspect discussed in the paper was that in the soft materials encountered in the Natal estuaries, cone penetration testing was very useful but there were limitations. As previously pointed out the system is essentially semi-empirical and relies on correlating cone pressures, or friction ratios, with other soil parameters so that locally applicable correlation factors can be established. The paper refers to field research conducted to correlate cone pressures with undrained shear strength measured by vane testing. In i I / DEPTH OF SUBSOIL Z m CONE PRESSURE Qc MPa =I ·OMF\:J =2'OMPo FILL HEIGHT (m) qc-M~R-a-t---:!--T--:--::::i:::::a~;:t:-;'---f----:i--'f"~S~ETTLEMENT(m) ctm I,!;i 2B NOTE: CONE PRESSURE LINES 4,2and I INDICATE T VAWES a --- I/mv = 5/2 (~ .. 3000) FINE TO MEDIUM SAND. b --- I/mv = 5/3 (ctc + 1500) CLAYEY SAND. EXAMPLE: H = 10m; 8 = 20; Z = 10m; qc = 1,0MPa ASSUME; ct.m = 1,5; y. = 20; '(5 = 15 kNI m3 : . SETTLEMENT = O,GOm CPT SETTLEMENT ESTIMATION CHART Fig. C·3·1 117 order to do this vane shear equipment was manufactured which could be used with the CPT rig. Figure C3.2 shows a diagram of the vane shear apparatus. It consisted of a retractable vane mounted in a nose cone attached to a string of CPT rods which had been modified to have square sockets at one end of each inner rod and matching square plugs at the other end. Two sizes of vanes were built to enable a range of shear strengths to be measured. A torque measuring spanner was manufactured for the head since commercially available torque wrenches had inadequate measuring sensitivity. The torque measuring system, which comprised a strain gauged bar connected to a chart recorder system, was calibrated in the laboratory. 100mm . x 50mm. 4 BLADE VANE RECESSES FOR VANE BLADES I I -=======-=--==~==::.=::: .. NOSE CONE Figure C3..2 : Vane shear apparatus SQUARE PLUG SQUARE SOCKET :E=~~+-~~ ~-:;.~-:. E ROO INNER E ROO OUTER Field testing was similar to cone penetration testing; the vane, the CPT casing and inner rods were pushed into the soil using the CPT rig. Penetration was stopped at 0,5 m intervals, the thrust transferred from the outer casing to the inner rods to push the vane out of the nose cone housing. The torque wrench was then inserted into the top rod socket and the vane rotated. Since the CPT rig prevented complete rotation of the torque bar a device was used which allowed an offset position for the bar. The initial position of the vane was carefully noted so that after rotation the outer casings could be advanced so that the vane finished in the protected position in the nose cone. Post peak residual shear strengths were easily measured and the sequence recorded on chart. A dummy vane comprising only the shaft without the vane blades was also made and pushed into the subsoil to measure a calibration zero for the system. The complete system worked extremely well and took only a little longer than conventional CPT measurements. It had advantages over measuring undrained shear 118 strengths by CPT since no soil dependent factors were involved. However the promise, or at least potential, for measuring other parameters, viz ~ with the CPT mitigated against further development of the vane equipment which was intended only for calibrating the CPT. Conventional CPT's were made in a soft clay at Umhlangane (Sea Cow Lake) and vane shear tests conducted at positions approximately 0,5 m away to allow direct comparisons of the two sets of data. These indicated that Nk = 18,4, if the overburden pressure term avo is ignored and Nk = 15,8 if it is included (Jones et aI, 1975) - Figure C3.3. The line shown is the linear regression through the origin and gives a correlation coefficient of 0,80 which is considered to be satisfactory. 80 70 60 a:: 60 \, Averages wL - wp mins Sea Cow 52 35 13 56-35 21 39 TSSH 23 19 58 40-18 22 3,6 TSPC 17 16 67 32-20 12 2,5 LPC - 7 93 NP 2,5 ? From Table C4.1 and Figures C4.15 and C4.16 it can be seen that there was a qualitative relationship between the consolidometer-cone t50 and the consolidometer t50 and between the former and the soil type. The derivation of a more closely defined relationship on the basis of the evidence would however, not be justified since there was a large gap in the data in the mid range. Nevertheless the correlation for the Sea Cow Lake clay (and it is the more clayey materials which give rise to embankment problems) was sufficiently close to warrant further investigation. 0 10 20 30 40 eN! ~ ~ ~O z 0 j: 60 u IAJ ..J "'- III 70 Q 80 90 100 0 0 10 20 30 eN! ~ 40 :z: .s::J ~ 50 j: U III ..J ~ 60 Q TO 80 90 100 0 x-x ------- x __ ~, x---...... x~ '\ ~ SILTY C A SEA COW \\ \ \ ~ ~ \ \ 10 100 TIME (MINS .) LABORATORY CONSOLIDOMETER TIME - SETTLEMENT TESTS FOR RANGE OF SOIL TYPES 10 TIME (MINS) 100 1000 FiO. C·4·15 1000 LABORATORY CONSOLIDOMETER-CONE TIME- SETTLEMENT . TESTS FOR RANGE OF SOIL TYPES FiO. C·4·16 140 As shown in Figure C4.1 the consolidometer-cone could be seen as a plate loading test approximately half the diameter of a conventional consolidometer. Whatever the shape of the volume of soil significantly stressed under the cone, ie sphere, hemisphere or cylinder, it is apparent that the drainage path length for consolidation is a function of the radius of the cone (17,5 mm). Since the consolidometer drainage path length for double drainage is 10 mm, it could be expected that consolidation times for the consolidometer-cone would be about triple those for the consolidometer. Also a significant part of the soil stressed, ie that next to the cone, has a much shorter drainage path length so that consolidation times of less than triple those for the consolidometer might be expected. For the clay sample the average values for the t50 and t90 times for both the consolidometer and the consolidometer-cone are given in Table C4.2. Table C4.2: Consolidometer and consolidometer-cone t50 and ~ times for Sea Cow Lake t50 mins t90 mins av t90/ t50 Consol Sample 7 10; 15; 13; 10; 11 70; 105; 65; 45; 45 5,52 Sample 8 21; 26; 17; 17; 13 100; 140; 90; 100; 65 5,26 Cone 24; 25; 26 74; 100; 150 4,5 37; 52; 60 180; 250; 290 I Cone/Consol 12,44 1 2,11 1 I The consolidometer results for the Sea Cow Lake clay in Table C4.2 are shown in Figure C4.14. Since the ratio of the t90 and t50 times should be the same as the ratio of the time factors Tv ie 0,848 to 0,197 = 4,30 this would suggest that the estimated t90 times for the consolidometer tests given in the Table are probably too high and include some secondary compression. Unquestionably the consolidation test results indicate a significant secondary compression component and because the test times were limited to 24 hours for each load increment, the estimation of ca:' the coefficient of secondary compression, and 141 hence t100 is relatively inaccurate. Selection of different slopes for Cat on the detailed laboratory test results gives a range of tw from 100 to 150 minutes and the corresponding range for t50 is 21 to 23 minutes, hence the t90/t50 changes from 4,7 to 6,5. Therefore although the average measured ratio in the Table is about 5,4 compared with the theoretical value of 4,30, this difference is not believed to be significant, since it is within the range of accuracy of the measurements from the consolidometer data. The corresponding ratio for the cone tests is 4,7 which suggests that the slope of the settlement - log time plots matches closely the slope for the theoretical one dimensional consolidation. Since consolidation around the cone must be three dimensional the ratio of Tv at a90 to that at aso would be expected to be somewhat higher; for example if Blight's (1968) solution for the consolidation of a sphere is considered applicable to the cone, then the ratio of Tv90 to Tv50 is about 1,35 to 0,21 viz 6,4; however, no significant inference was drawn from this anomaly. For the clay the cone tests give t50 and tw about 2,4 and 2,1 times longer than those for the consolidometers - Table C4.2. Since the time factors, Tv for the consolidometer T and the cone T as a sphere are vo vc approximately the same at tso (0,197 and 0,21) then from: T a2 C =-"- 'f t I then a = (2.4 x 0.197 Xl(0)~ 0.21 • 15 nun C4.1 C4.2 ie the equivalent sphere has a radius 15 mm, a little less than the radius of its generating cone, 17,5 mm. 142 This analysis demonstrates that the volume of soil consolidating under the cone is about the same as a sphere of the same radius as the cone. Since the cone t50 and t90 times are about two and a half times longer than the equivalent consolidometer times then : The foregoing results were very encouraging and tended to confirm that the cone in the constant stress mode could be used to measure the consolidation properties of clay. However when the same approach was adopted for the tests on the other soils a less encouraging picture emerged. The soil type TSSH results are examined first because this series was conducted in the Prototype 3 apparatus which was significantly superior to the previous prototypes. The relevant test results are given in Table C4.3. Table C4.3 : Consolidometer and consolidometer-cone t50 and '90 times for TSSH Sample Depth Consol Cone Run Load No mm t50 (min) No t90 (min) t50 (min) t90 (min) 1 0,19 0,68 1,5 17 1 1 2 90 0,17 0,94 - - - - 3 150 0,17 0,66 20 30 2 2 4 190 0,17 0,62 20 80 3 3 5 210 0,20 1,04 4,0 17 4 7 6 250 0,18 0,62 6,0 80 5 11 0,18 0,77 3,6 36 I t9O/t50 I 4,3 I 10 I cone/consol t50 - 20 t90 = 47 143 It is apparent that the various ratios given in the Table are very different from those for the clay samples (Table C4.2). The consolidometer showed very short times for settlement and in almost all cases the t90 times were less than 1 minute. Since measurements had only been started at 0,5 min the estimates of t50 can only be considered as very approximate, nevertheless the t90 : t50 ratio of 4,3 is the same as the theoretical ratio for one dimensional consolidation and suggests that the samples have undergone consolidation. For the consolidometer-cone the corresponding t90 : t50 is 10. Re-examination of the cone plots, Figure C4.12, however, shows that some of the results exhibit significant secondary compression or creep and this has a marked influence on the t90 : t50' The cone results were modified to determine a revised t100 and the t50 and t90 times are given in Table C4.4. Table C4.4 : Modified consolidometer-cone t50 and ~ times for TSSH t50 min t90 min Run No 1 2 3 4 5 1 2 3 4 5 Load 1 2 3 7 11 1 2 3 7 11 Original 1,5 3,0 20 4,0 6,0 17 30 80 17 80 Revised 1,2 1,9 17 3,6 4,5 9 13 80 15 20 Average t 2,8 (excl Run 3) 14,25 (excl Run 3) I t90/t50 I 7,5 I 6,8 I 47 I 4,2 I 4,4 I Average 5,5 I It can be seen that the revision makes a considerable difference particularly to the cone t90 and hence the ratios in Table C4.3 should be modified. The cone t90 : t50 becomes 5,5 (excl Run 3); the cone: consol t50 ratio becomes 16 and the t90 ratio becomes 18. The modification to the cone results therefore leads to consistent ratios for the TSSH soil, but despite this improvement these ratios are dissimilar to those for the Sea Cow Lake samples. These are compared in Table C4.5. I .- - 144 Table C4.5 : Comparison of cone / consolidometer ratios for tso and 40 for TSSH and Sea Cow Lake ~========~===========, Soil Cone/Consolidometer Ratios tso Sea Cow Lake 2,4 2,1 TSSH 16 18 It was argued previously that the cone/consolidometer Sea Cow Lake soil ratios of about 2 - 2,5 were what might be expected from the relative drainage path lengths for the consolidometer-cone and consolidometer viz 17,5 mm and 10 mm. The TSSH soil gave very different values and by the same line of argument as before the implied drainage path length for the cone was: 1 a = (16 x 0,197 x 100)2 0,21 39 mm There was no obvious explanation for this discrepancy; the stress levels in both the Sea Cow Lake and the TSSH series of tests were similar, the former loads being in the range 21/2 - 7 lbs and the latter 1 - lllbs. The Sea Cow Lake tests showed no correlation between stress and tso whereas the TSSH showed a consistent, although non linear, relationship - Figure C4.17. Tests on the latter also showed - Figure C4.18 - that the required cone pressure qc appeared to be a function of the test depth and a similar, but somewhat less marked, trend was apparent for the qc and void ratio relationship. However the initial voids ratios for the TSSH samples were also a function of depth - Figure C4.19 - since the samples were prepared in compacted layers in the container. It would be expected that for this laboratory prepared clayey silty sand the permeability would be directly related to the initial void ratio. This is shown in the laboratory consolidometer test results when void ratio is plotted against tso, Figure C4.20. o SEA COW I!!O mini . 20 12 40 ~o LOAD AGAI NST \50 FOR )I. SEA COW AND TSSH. 10 / 1 0 )I. . 0 ~ . 0 I 0 c 3 4 0 2 ./J. 0 /11 2 4 6 X TSSH '50 .. Inl. Fig. C ·4·17 o INITIAL VOID RATIO 1,0 O,II!! o,eo 0.15 !!O SAMPLE TSSH: .. IL AGAINST VOID RATIO .. 'l,c 40 u I I O,8!! i I 0,80 I I 2 3 4 , '!!O mine. I Fig. C'4'20 146 Detailed consideration of test series, LPC and TSPC, show that the results were similar to those for the TSSH series. The LPC was a non plastic silty sand which had initial laboratory void ratios in the range 0,48 _ 0,62; the consolidometer tso were in the range of about 0,10 - 0,20 minutes and t90 from 0,5 - 1,0 minutes. The cone tso were in the range of about 1 - 21/2 minutes and t90 in the range 5 - 10 minutes. The consolidometer-cone testing equipment was in the initial stages for the LPC series and was relatively crude, nevertheless the results are similar to those for TSSH. The required load to induce controlled settlement appeared to be a function of sample depth but from the data it could not be ascertained whether the tso or ~ times were also dependent on depth. Sample TSPC was a slightly clayey silty sand. The initial laboratory void ratios were in the range 0,72 - 1,07 and were a function of depth. The consolidometer tso were in the range of about 0,2 - 0,5 minutes and the t90 from about 0,5 - 2 minutes. The consolidometer-cone tso were in the range 0,5 - 9 minutes with an average of 2,5 minutes and the t90 in the range 4 - 50 minutes with an average of 20 minutes. The latter was significantly reduced by correcting the plots for excessive creep as was done for the TSSH results, resulting in an average cone tso of 2 minutes and t90 of 12 minutes. The LPC and TSPC results are similar to those for the TSSH series and the more pertinent ratios are given in Table C4.6. Table C4.6: Consolidometer and consolidometer-cone times and ratios for all samples Test Average Time (Minutes) LPC TSPC TSSH Sea Cow Consol tso 0,15 0,2 0,18 15 t90 0,75 0,80 0,77 81 t90/tSO 5,0 4,0 4,3 5,4 Cone tso 1,5 2,0 2,8 37 t90 .7,0 12 14,2 174 t90/tSO 4,7 6,0 5,1 4,6 Cone/consol tso 10 10 16 2,4 Cone/consol t90 9 15 18 2,1 147 The overall picture from the comparison of the results of the silty and clayey sands with those of the Sea Cow Lake clay is that the clay results appear to be consistent with the idea of the cone acting as a form of consolidometer and thus measuring consolidation characteristics, whereas in the sandier materials the same simplistic model does not apply since the cone/consolidometer time ratios are much higher than the expected value of about 2 to 2,5. Re-examination of the results showed that although the applied pressures were similar for all the tests, the resulting strains were very different. Typical measured chart displacements together with the actual L VDT settlements are given in Table C4.7. Table C4.7 : Measured strains for consolidometer-cone tests Sample No of Chart Distance mm Cone Settlement Runs Pressure Cone Range Average kPa mm LPC 8 97-280 170 40 10,6 TSPC 21 23-224 88 55 11,0 TSSH 5 133-288 229 20 14,3 Sea Cow 7 12-49 27 21 2,7 The measured consolidometer settlements for the pressure range equivalent to the cone test pressures; the coefficients of volume compressibility, Il\, (in the pressure range 50 150 kPa); the compression indices Cc' (in the 50 - 100 kPa range) and the initial void ratios are given in Table C4.8. Table C4.8 : Consolidometer test results Sample Initial Void • 6H Il\ Cc Ratio MN/m2 mm LPC 1 + 2 0,55 0,08 0,16 0,025 LPC3 0,60 0,11 0,21 0,034 TSPC 0,88 0,35 0,65 0,16 TSSH 0,97 0,57 0,75 0,20 Sea Cow 2,72 0,61 0,6 0,50 * for e ulvalent q p ressure ra nge to cone tests 148 A comparison of the settlements in the consolidometer tests, for the pressure range equivalent to the cone tests, with the settlements for the corresponding consolidometer cone tests is shown in Table C4.9 as the ratio of consolidometer-cone to consolidometer settlements. Table C4.9 : Cone : consolidometer settlement ratios Material Settlement Ratio Cone/Consol LPC 66,2 TSPC 16,9 TSSH 19,1 Sea Cow 4,3 It was clear that the cone settlements in the more sandy materials were very much larger than the equivalent consolidometer settlements, whereas for the clay the ratio was about 4,3. The latter was reasonable if the cone was considered to influence a sphere of approximately the same diameter as the cone ie 35 mm compared to the consolidometer sample thickness of 19 mm and that the lateral constraint conditions for the cone and consolidometer were different. For all the other samples, however, the relative strains were inordinately high and there could be only one explanation. It was that for these materials the consolidometer-cone test was not measuring consolidation. The erratic nature of many of the earlier tests in particular had suggested that a failure situation was being developed. With the improved equipment, however, the results for the TSSH soil were consistent and during testing gave rise to the hope that they could be usefully interpreted in terms of consolidation. To paraphrase the words of the French general witnessing the Charge of the Light Brigade, "C'est magnifique, mais ce n'est pas la consolidation (guerre )". It is worthwhile giving some consideration to these results; they must represent a bearing capacity failure of a buried foundation and Figure C4.18 shows the ultimate bearing pressures against depth, as well as against initial void ratios of samples subsequently taken at or close to the cone test depths. Figure C4.19 shows the correlation of void ratio with depth the form of which is to be expected since the 149 sample was compacted in layers and it is apparent that the governing factor for the bearing capacity is not only the depth, but also the initial void ratio of the sample. In other words cone penetration is increasingly resisted by the increasing in situ stresses with depth and by the increasing soil density with depth, and the overall result is a decreasing rate of failure. The actual cone movements in these materials is about 10 - 15 mm which is very large compared with the diameter of the cone, viz 35 mm. Although this re-appraisal of the proposed consolidometer-cone system produced mixed conclusions, namely that in sands no useful results could be obtained, but in clays the results were encouraging, the latter justified the continuation of the work. The ultimate purpose was for in situ testing of clays in which the times for embankment consolidation were expected to be large. In such materials the values of the coefficients of consolidation, cy , at which practical problems of slow settlements would be expected to start becoming significant, would be in the range of about 5 - 20 m2/year. The sandy materials which were tested in the laboratory have coefficients of consolidation of about 50 m2/year and the Sea Cow Lake clay has a cy of about 0,5 - 1 m2/year. The range of materials selected for the laboratory programme was too wide and intermediate materials were required. The results, however, emphasized the more significant problem of whether or not the cone test was measuring consolidation settlement or shear strain, or a variable combination of both. In order to test this the author conceived the idea of installing a piezometer in the cone to measure the rate of dissipation of pore pressure during settlement of the cone. The following section, C5, describes the laboratory piezometer cone. 150 C5 lABORATORY PIEZOMEfER CONE The laboratory piezometer cone was designed by the author in early 1975 to incorporate a pore pressure measuring system into a standard mechanical cone which was being used for the laboratory consolidometer-cone testing. The previous consolidometer-cone used the standard friction sleeve Dutch mechanical cone modified only to the extent that the inner rod had been replaced with a longer rod to accommodate the weight platform. The standard cone, however, was extensively modified to fit in the piezometer and this is shown in Figure C5.1. Fortuitously the pressure transducers which were available were eminently suitable for the purpose and the same type continues to be used in piezometer cones fifteen years later. The particular transducer used, Kyowa PS - 1 KA had a pressure rating of 1 bar or 100 kPa. The transducer was a thin disc of 5 mm diameter and hence could very conveniently be fitted into the cone. The filter elements were set into the cone face as shown. The four elements consisted of ground to shape ceramic sections from conventional consolidometer drainage discs. The pressure transducer sensitivity was 1 m V IV at full load (loa kPa) and the excitation voltage was 3 volt. The output was led directly to a chart recorder and resulted in the full chart width, 250 mm, representing a pressure change at 1 m V chart setting of 0,03 m V for 1 kPa or 1 mm representing 0,13 kPa. The results of the first test in a soil (TSPC) are shown in Figure C5.2. The shape of the cone settlement curve appeared to be a good example of the previous series of similar cone tests, but that of the pore pressure curve was puzzling. It showed an increase in pore pressure from the initial zero value for 4 minutes until a peak was reached at a pressure of 1,82 kPa. The cone pressure for this test was 30,8 kPa (7 lbs load) so the maximum recorded pore pressure was only about 5% of the cone pressure. The pore pressure subsequently dropped very rapidly, but then, even after a long time, did not essentially change and remained higher than the initial value. Experience of operating the consolidometer-cone in sand had led to the procedure of adding the load in increments until sufficient was in place to promote a significant settlement. For this particular test the initial 4 lb load was supplemented by a further 2 lb then a further 1 lb within about one minute. 60° CONE WITH FOUR FILTERS ~ ::> '0 0 Q ~ a ~ :.: <0 0 /0 20 30 40 !l0 60 TO 80 90 100 / 0 TRANSDUCER SLEEVE THREADED ONTO TRANSDUCER BLOCK SLIDING ON BOSS BLOCK THREADED INTO CONE. CENTRAL HOLE FOR CONNECTIONS . ACTUATING ROD WITH GROOVE FOR CABLE THREADED INTO TRANSDUCER BLOCK SLIDING THROUGH BOSS. / LABORATORY PIEZOMETER CONE Fig. C·5·' ·r----..... ---... .... / " , /0 TIME MNUTES TSPC - H • TSPC - U - x- - TSPC- U Adjusted - -----. SEA COW - H - . .......0-. - SEA COW - U H' SETTLEMENT U' PORE PRESSURE /00 /000 LABORATORY PIEZOMETER CONSOLI DOMETER-CCf\E TIME- SETTLEMENT AND TIME- PORE PRESSURE DISSIPATION FOR TSPC AND SEA COW LAKE . Fig. C·5· 2 152 Despite this delay in applying the full load there was still a significant lag before the peak pressure was registered. This was believed to be due to the measuring system itself being relatively soft, ie large volume, and lack of any de-airing procedures. The peak pore pressure recorded was very much lower than the cone total pressure (viz 5%) and this low response was assumed to be because the peak pore pressure under the cone had already largely dissipated in the time taken for the instrument to respond. The subsequent rate of dissipation was masked because of the lag in response; if the lag is assumed to be constant after the peak at 4 minutes, then correction should be possible by shifting the dissipation curve by 4 minutes to the adjusted position shown. Although the total time for 90% dissipation was about 20 minutes, the same as that for the cone to travel 90% of its total settlement, (Figure C5.2) the shape of the pore pressure dissipation curve was different from both the laboratory consolidometer-cone and consolidometer settlements Figures C4.1S and C4.16. This, however, was not unexpected because as already discussed in Section C4 it had been shown that in the silty sands the settlement measured and hence the pore pressure dissipation measured was not due to consolidation. The pore pressure dissipation curve for sample TSPC in Figure CS.2 did not return to the original zero and this was because the depth of the cone had increased by, in this test, about 20 mm. Since the increase in pressure in the saturated sample was hydrostatic, it would have been expected to be 0,2 kPa which was very close to that measured. It was concluded from this initial testing with the piezometer cone that the instrument itself required improvement, together with the de-airing procedure to ensure that the response time was minimized, but that the system was satisfactory. Modifications were made to the cone to decrease the volume around the transducer to about one third of the original volume and shaped to minimize the possibility of trapping air bubbles. The holes from each of the filter element recesses were made finer to minimize the volume of the system. A method of de-airing and calibrating the piezometer cone was devised as shown in Figure CS.3. The perspex block was clamped over the cone and tightened in place with clamping screw; the water leads were connected to the laboratory triaxial equipment, which used de-aired water, and a 153 pressure difference set up to cause flow through the system. After allowing an initial period to de-air, calibration of the system was accomplished by setting a series of pressures and measuring the corresponding chart displacements. It was observed that at changes of pressure the piezometer cone response was very rapid. SEALING RI NG Figure C5.3 : De-airing and calibration adaptor YOKE TO BACK OF CONE CLAMPING SCREW Few such tests were carried out and the records of these are only available in fragmentary form. A typical pore pressure record for the Sea Cow Lake clay is shown in Figure CS.2 together with the simultaneous cone settlement record. In order to minimize initial loading increment problems the selected load was placed as one load. In some cases this induced failure and these results were discarded after cursory inspection. Where an apparent consolidation type settlement took place the pore pressure were generally as shown in Figure CS.2; some however showed erratic fluctuations of pore pressure. Subsequent examination of the large block sample showed open fissures and small sandy inclusions and it was surmised that these had caused the pore pressure variations. The typical result shown in Figure CS.2 for the Sea Cow Lake clay confirmed that the rate of settlement was closely matched by the rate of excess pore pressure dissipation, and therefore that the measured settlement was predominantly primary consolidation. 154 There was still some initial lag in the pore pressure response but subsequently dissipation stopped before completion of settlement. The maximum pore pressures were about 10% - 20% of the cone pressures and in some cases a residual pressure would continue which could not be explained by the change in depth since this was only about 3 mm. The pore pressures however, were very low, say up to 5 kPa, compared to full load of 100 kPa and the small residual pressure was assumed to be due to the system. The overall conclusion was that in clays in the laboratory the consolidometer-cone was shown by the piezometer results to satisfactorily model consolidation. It was therefore necessary to test the consolidometer-cone (without piezometer) in the field. During this period, May - June 1975, the opportunity arose to carry out cone penetration testing along a section of the proposed alignment for the National Road at Umkomaas, South Coast Natal. The National Institute for Road Research 100 kN Goudsche Machine Fabriek rig was used, with the modified strain gauge load measuring system described in C3.3. The standard mechanical friction sleeve penetrometer was used in the consolidometer-cone mode. The approximate position of the proposed alignment then closely followed the coast and involved an embankment about 800 m long across a low lying swampy area. The mechanical cone penetration test results indicated extremely poor materials and it was obvious that embankment construction would have been difficult and the future performance would have been poor. As a result the proposed road was subsequently realigned to the present route, which is further inland. During this investigation the field version of the consolidometer-cone was tried for the first time. The equipment is shown in Figure CS.4. It consisted of entirely standard cones and rods except for a longer than standard final inner rod to which the weight platform was fixed. The procedure was that a normal mechanical CPT was carried out to a suitable clay layer depth judged from the cone and friction sleeve measurements. The upper inner rod was removed and replaced with the special rod and load platform. A L VDT was positioned and connected to the chart recorder. Loads were added to the load platform until settlement was induced and then the time-settlement data recorded on the chart. The results were dissapointing; the dividing line between insufficient FIELD CONSOLIDOMETER - CONE TESTING SHOWING LOAD PLATFORM, SETTLEMENT MEASURING LVDT TO CHART RECORDER. Fig. C· 5 · 4 156 weight to cause significant movement and sufficient to cause failure was so fine that very few consolidation curves were obtained. A large amount of weight was necessary (over 100 lb) even in a soft clay; this was expected, since cone pressures in this material were about 500 - 700 kPa - ie undrained shear strength of about 30 to 40 kPa - which is equivalent to a load of about 150 lbs and a significant proportion of this was expected to be necessary to cause settlements. The amount of weight was not in itself a major problem; the critical issue was that a relatively small increment could cause the cone to plunge down to the full extent of its travel. The loads and platform were removed, the cone advanced to a new position and a further test attempted. The success rate was low and the time taken for full settlement was long, being two or three hours and it became clear that the prognosis for the consolidometer-cone test as a routine field procedure was limited. The operation, however, of the chart recorder and strain gauge load cell with the mechanical cone proved to be very successful in the field. The enhanced accuracy and convenience compared with the standard pressure gauges and manual recording of large numbers of readings was readily apparent. These experiences overcame the initial reluctance to operate this type of equipment in the field, bearing in mind both the rugged conditions and the relatively low level of sophistica~ion of both operators and equipment then current in site investigation in South Africa. Further important factors were that the success of the laboratory cone piezometer, the earlier experience of designing and building load cells to measure cone pressures, and the confidence of operating such equipment in the field had all led to the idea of building a field electric piezometer cone. The purpose of this was to measure in situ pore pressure dissipations and hence directly obtain the consolidation characteristics of the alluvial deposits, which was one of the ultimate objectives of the research. The development of the field piezometer cone is described in Section C6. In summary, at the end of the research on the use of the consolidometer-cone ie 1977 , the overall standard of cone penetration testing in South Africa had been raised very greatly from that reported by Webb (1974) : • • • 157 the standard Dutch mechanical cone with friction sleeve was in common use, hydraulic load cells with high and low pressure gauges operating independently of the penetrometer rig hydraulic ram were mandatory, appreciation of the value of the results obtained from cone penetration testing had been significantly raised. The author was instrumental in achieving these through advocating the technique and the improvement of the equipment during this period in consulting engineering, at the Roads Department of the Natal Provincial Administration and later at the CSIR. The latter not only allowed the research and general development of the equipment to be conducted, but through discussions, lectures and publications (Jones, 1974), and close ties with the road authorities, the awareness and knowledge of the technique was extended. By 1974, it was clear that the standard cone load measuring equipment lacked the sensitivity for meaningful results to be obtained in the very soft alluvial materials, although the CPT did at least give a very definite indication of the presence of such materials. The author, nevertheless, was convinced that the inherent advantages of cone penetration testing, viz, rapid and relatively inexpensive testing of deep multilayered materials, should be exploited. A statement by Heijnen (1974) on penetration testing in the Netherlands was not considered by the author to be universally applicable, "Very seldom the cone resistance values are used for the prediction of settlements of compressible soil layers. As to be expected the employed relation is not very reliable". The author's view was that not only was the cone penetration test suitable for the prediction of settlements, but that the research then being conducted would enable not only settlements to be adequately predicted on the softer materials, but would also allow satisfactory estimates of consolidation times to be made. 158 C6 DEVEWPMENT OF SOUfH AFRICAN FIELD PIEZOMETER CONE Part C5 described the laboratory piezometer cone which had been developed by the author to determine whether the laboratory consolidometer-cone settlements conincided with pore pressure dissipation and hence measured consolidation settlement. It was concluded that in the clay tested the rate of dissipation of excess pore pressure was similar to the rate of settlement and therefore that the settlement was due to consolidation. The experience of using a piezometer in a cone had led to the belief that direct measuring of pore pressure dissipation was the preferred route for estimating consolida tion characteristics. The laboratory mechanical cone fitted with a piezometer had shown a number of aspects of the pore pressure response to loading in a clay; viz : • the pore pressure increased immediately on loading • at constant load (ie in the consolidometer-cone mode) the pore pressures gradually decreased • the pore pressure increased immediately on loading but instantaneously decreased on unloading to some intermediate value • if the cone was clamped in place before unloading the instantaneous pore pressure decrease was avoided and dissipation continued. What was particularly interesting, was that the pore pressure dissipation did not require ongoing penetration. It was about this time, mid 1976, that the 1975 papers by Torstensson and Wissa became available. These addressed the rate of penetration around piezometers and together with the author's experience were sufficient to confirm that a cone for simultaneously measuring cone and pore pressures was a necessary development. Initially the author considered measuring cone loads through the standard mechanical system of inner rods to the surface, and measuring pore pressures as in the laboratory piezometer cone. This was abandoned because of the problem of accommodating both inner rods and a cable within the conventional sounding casings which had an internal diameter of 16 mm, with the inner rods having a diameter of 15 mm, hence there was no space for a cable. 159 Some thought was given both to threading a continuous cable through the inner rods and to having a discontinuous cable through the inner rods with connectors at each end of the rods. Both of these would have been difficult; the former would have seriously weakened the inner rods - but was a possibility if the equipment was only to be used in softer materials - and the latter would have necessitated a large number of waterproof connectors which would have been a major source of problems. If the inner rods were drilled to accommodate cables then in dense materials rod pressures of 200 MPa would result. This is not an excessive pressure for high quality steel rods for simple compression and the reduction in area would not be intolerable even at full load. In practice, however, the inner rods, unless the ends are very carefully machined and maintained in perfect alignment with one another, will be subject to high eccentric loads and buckling and overstress of the ends would inevitably result. Despite these potential problems the proposed system had some benefits, provided some restriction on the cone loads could be maintained. The obvious benefit would have been the use of the existing mechanical cone systems, together with the strain gauge load cells for measuring cone and friction sleeve loads. The alternative was to dispense with the mechanical cone system altogether and change to cones incorporating load cells and hence eliminate the inner push rods. Electrical cones had been in use elsewhere for some years, but not in South Africa, and de Ruiter's 1971 ASCE paper illustrated such strain gauge penetrometers - Figure B2.S. PLUGGED SLOT FOR CONE REMOVAL KEY AND ACCESS TO DE AIRING HOLE FILTER Figure C6.1 : Field piezometer cone STRAIN GAUGE CONE LOAD CELL . : , LITHOLOGY LlTOLOGIE Q Q Alluvium, sand, calcrete Qb Aeolianite, sand, clay, limestone Alluvium, sand, kalkreet EOlianlet, sand, klei, kalksteen - T-Qa limestone, clay, conglomerate T-Qb Limestone, sandstone, conglomerate T-Qk Sand, limestone T-Qn Aeolianite, dune sand Kalksteen, klei, konglomeraat Kalksteen, sandsteen, konglomeraat Sand, kalksteen Eolianiet, duinsand T Tg Silcrete Siltstone, limestone, calcarenite Silkreet Tu Sliksteen, kalksteen, kalkareniet Ki limestone Kmh Conglomerate, sandstone Kmz Limestone, clay Ksu Breccia, tuff, trachyte, melilite basalt, carbo~tite Kalksteen Konglomeraat, sandsteen Kalksteen, klei Breksie, tuf, tragiet, melilietbasalt, karbonatlet K Sandstone, conglomerate, marl Conglomerate, sandstone Sandstone, mudstone, shale Siltstone, sandstone, conglomerate Kma Sandsteen, konglomeraat, merrel Kmg Konglomeraat, sandsteen Ks Sandsteen, moddersteen, skalie Kz Sliksteen, sandsteen, konglomeraat - J-K Mudstone, sandstone, conglomerate Moddersteen, sandsteen, konglomeraat Jb Rhyolite, syenite, basalt, tuff, breccia, conglomerate, sandstone Jdr Basalt Jj Rhyolite Jm Basalt Js Basalt, tuff, breccia J Rioliet, sieniet, basalt, tuf, breksie, konglomeraat, sandsteen Basalt Rioliet Basalt Basalt, tuf, breksie Jd Dolerite Ie Conglomerate, sandstone II Basalt Jp Breccia, agglomerate, tuff JI Granophyre Doleriet Konglomeraat, sandsteen Basalt Breksie, agglomeraat, tuf Granofier lib Mudstone Rm Sandstone, mudstone, shale liny Mudstone, sandstone Moddersteen Sandsteen, moddersteen, skalie Moddersteen, sandsteen RC Sandstone, siltstone limc See RC, Re, Rm lit Mudstone, sandstone k Sands teen, sliksteen Kyk lie,lie, lim Moddersteen, sandsteen - lie Mudstone, sa~dstone lint Sandstone Moddersteen, sands teen Sandsteen '-- P- li Shale, sandstone, mudstone, coal P- lii Mudstone, sandstone P- lisk Shale, mudstone, sandstone Skalie, sandsteen, moddersteen, steenkool Moddersteen, sandsteen Skalie, moddersteen, sandsteen Pa Mudstone, sandstone Pf Shale Pp Shale Ps Shale, sandstone Pw Carbonaceous shale I' Moddersteen, sands teen Skalie Skalie Skalie, sands teen Koolstofhoudende skalie Pc Sandstone, shale Pk Shale Ppr Shale Pt Shale Pwa Sandstone, shale p Sandsteen, skalie Skalie Skalie Skalie Sandsteen, skalie Pe Shale Pko Sandstone, shale Ppw See Ppr, Pw Pv Sandstone, shale, coal Skalie Sandsteen, skalie Kyk Ppr, Pw Sandsteen, skalie, steenkool Pem Mudstone, shale, sandstone Pm Sandstone, shale, coal Pr Sandstone, shale Pvo Shale Moddersteen, skalie, sandsteen Sandsteen, skalie, steenkool Sandsteen, skalie Skalie - C- Pd Tillite, sandstone, mudstone, shale C Tilliet, sandsteen , moddersteen, skalie Db Shale Dc Shale, sandstone Dt Shale, siltstone, sandstone Skalie Skalie, sandsteen Skalie, sliksteen, sandsteen 0 Obi Shale, siltstone, sandstone 01 Shale, sandstone, diamictite Ow Quartzitic sandstone, shale Skalie, sliksteen, sandsteen Skalie, sandsteen, diamiktiet Kwartsitiese sandsteen, skalie S Sn Quartzitic sandstone, shale, tillite Kwartsitiese sandsteen, skalie, tilliet f-- O- S Quartzitic sandstone, arkose, shale Kwartsitiese sandsteen, arkose, skalie 0 Op Quartzitic sandstone, shale Ope Quartzitic sandstone Kwartsitiese sandsteen, skalie Kwartsitiese sandsteen -€ £k Sandstone, conglomerate, shale £1 Granite, syenite, foyaite Sandsteen, konglomeraat, skalie Graniet, sieniet, foyaret - N- £ Biotite granite N- £k Biotite granite Biotietgraniet Biotietgraniet - Nnt Conglomerate, mudstone, limestone, schist Konglomeraat, moddersteen, kalksteen, skis N Ntu Amphibolite, gneiss, schist Amfiboliet, gneis, skis Nmp Gneiss, granulite Gneis, granuliet I' SCALE: ........................... I: IlXJJUJL LITHOLOGY FIGURE A.3.lc '- ~ 160 The first prototype electrical piezometer cone was designed by the author and built in 1976; design drawing of the piezometer cone is reproduced in Figure C6.1. The technology was developed from experience with the laboratory piezometer cone and from the earlier strain gauge load cells systems which had been used both for vane shear testing and for the enhanced cone load measuring systems for the mechanical cone testing. The piezometer cone was calibrated in a modified triaxial cell - with the top plate of the cell having a 36 mm diameter hole and sealing rings - and tested in the cone penetration loading frame previously described - Figure C4.8. The combined cone load and pore water measuring penetrometer, called the piezometer cone or piezocone, performed extremely well in the laboratory trials and both cone loads and pore pressures could be accurately calibrated through the triaxial test equipment. The pore pressure system worked almost perfectly. The resolution at the full range of pressures tested was more than adequate. The pressure transducer was upgraded to the Kyowa PS - 2 KA (200 kPa), instead of the 100 kPa transducer used in the initial laboratory cone, despite the fact that even the upgraded transducer would limit the depth capability in soils giving a high pore pressure response. The response time of the pore pressure system was again noted to be variable and dependent on the efficiency of the initial de-airing and subsequent maintenance of saturation. These were not difficult to achieve and gave no cause for concern that there would be practical field problems. The four flush mounted face filter elements appeared to perform satisfactorily, but were difficult to manufacture. The purpose of face mounted filters was that it was assumed that the maximum shear and compressive stresses, and hence pore pressure development, would be at the face of the cone, thus filters half way up the cone would represent an average position. However, from experience with various piezometers, the idea was conceived of having a cylindrical filter element at the base of the cone where it would be less subject to damage during penetration than face or tip mounted filters. It was appreciated that a different pore pressure regime wou,ld be generated in this zone compared to the cone face, but since the primary purpose was to generate an excess pressure in order to measure its rate of dissipation, the magnitude of excess pore pressure was largely irrelevant, provided it was sufficient to sensibly measure. The practical simplicity of the cone base position made it the preferred situation. A cone was therefore made in this configuration. The filter element, as in the laboratory piezometer consolidometer-cone, - Figure CS.l - comprised ceramic filters used in conventional consolidation tests. Figure C6.2 shows the two cone types. FACE (Left) AND BASE (Right) FILTER CONES (LATTER WITH HOLE AND PLUG FOR UNSCREWING BAR . Flg.C·6·2(o) FRICTION SLEEVE PIEZOMETER CONE. OPTICAL ENCODER WITH RANGE OF PULLEYS. Fig,C·6·3 Fig. C·6·7 162 The load cell system was designed by the author on the basis of the knowledge gained from previous load cells made for measuring cone loads through the mechanical cone system inner rods. It was found that there was more than adequate space within the penetrometer body to provide sufficient cross sectional area for the load cell, so the design decision was to select the desired load range and provide a suitable cross section area. A solid cylindrical section was chosen since this was relatively easy to manufacture; flats were machined on the cylindrical section to ease the problem of cementing the strain gauges - Figure C6.3. The diameter of the central measuring part of the load cell was 24 mm, which after machining the four flats gave a cross sectional area of 425 mm2. This allowed a cone load of 85 kN at a steel stress of 200 MPa; with a modulus of 200 GPa the strain at full load would therefore be about 100 micro strain. The electrical strain gauge load cell design allowed for drilling out a central hole to decrease the cross sectional area and hence produce a range of load cells for designed pressure ranges with the same external sizes and fitting. The original strain gauge system consisted of a simple uncompensated bridge which was later (1977) improved to comprise four 90 degree, 120 ohm Kyowa KFC-2-D15-11 strain gauge rosettes in one fully compensating bridge to measure axial loads only and compensating for eccentric loads and temperature changes. With this bridge, and using a 5 volt excitation voltage, the maximum output from the load cell was 10 m V. In practice the cone pressures in different materials have a very wide range, varying from about 30 kN for dense sands to about 0,2 kN for soft clays; in addition it was possible by hitting rock or boulders, to develop instantaneous cone loads of about 60 kN before penetration was stopped. The first load cell was therefore a deliberate compromise. The soft clay load of 0,2 kN (equivalent to undrained shear strength of 15 kPa) produced an electrical output of only 0,02 m V which on a 100 division chart at 1 m V full scale gives only 2 divisions. The resolution in terms of undrained shear strength for soft clays was little better than say ± 3 kPa and if the system operated only moderately well this accuracy could be expected to be lower and would be inadequate for the measurement of undrained shear strength of soft clays. The system was, however, experimental and the primary purpose was to check the pore pressure measuring system in the field; the cone load system was deliberately detuned 163 to avoid overloading damage. Nevertheless, even at this detuned level, the sensitivity of the cone load measuring system was considerably superior to the then current standard of mechanical cones. The laboratory tests carried out in a triaxial test frame revealed a number of mechanical and electrical problems. Water leaked under pressure, despite the cone sealing ring, into the load cell space and caused electrical problems : under repeated loading some shift in the zero load readings was observed, and at the amplifications necessary to give reasonable chart output a considerable extraneous interference or noise was apparent. The pore pressure measuring system operated without problem. At this stage, the beginning of 1977, the author left the NITRR and joined the consulting engineering practice of van Niekerk, Kleyn and Edwards (VKE), who, at the time were designing a major national freeway project along the Natal South Coast which involved a number of crossings of rivers and flood plains. The investigations for these and subsequent monitoring is described in Part D. Fortunately, because of the cooperation between the Department of Transport, NITRR and VKE, it was arranged that the research equipment be transferred to the VKE soils laboratory so that the development could continue. Despite the pressures of the new post, some progress was made with the equipment during 1977. The strain gauge bridge was improved to include eccentric load and temperature compensation; techniques for cementing and waterproofing the strain gauges were developed and the sealing ring and cable gland were improved. The consolidometer-cone frame was utilized for testing prepared samples. These were clayey layers of different thicknesses in silty sands. The tests showed that thin layers, ± 10 mm, could readily be detected by the very marked changes in pore pressure response. Significant pore pressures were recorded in the clay layers and dissipations of these with time were observed. Field tests were conducted in the Pretoria area where softer materials could be found. The systems worked as had been shown by the laboratory tests and the field work was primarily to develop the operating techniques needed to deal with cables through the pushing devices and through the CPT outer rods. The instrumented cone had a short length of 8 core cable emerging from the 164 penetrometer which was connected to the main cable by a threaded plug and socket having eight connectors. This connection gave incessant problems and it proved impossible locally to find a suitable connector that was watertight and could cope with both the tension and bending at the plug and also fit inside the 16 mm internal diameter casings. The connector system was somewhat reluctantly abandoned in favour of having the penetrometer semi permanently attached to a 30 m length of cable. The reluctance was because damage to the cable would entail opening the cone in order to remove the cable. In fact it transpired that opening the cone was more frequently required to repair the cone rather than the cable and until recently subsequent models continued with the integral cable method. In mid 1978 the first commercial site investigation project using the piezometer cone was undertaken at Bafokeng Impala Platinum mine some 100 km west of Pretoria. The purpose was to establish the hydrostatic conditions within an existing tailings dam, which was then about 25 m high, so that decisions could be made regarding the stability and hence the potential safety for raising the dam. The piezocone was intended to perform as a piezometer by penetrating to selected depths and measuring the ambient pore pressures, but, in addition, it had the advantage of measuring cone resistances. The expectations were over-modest: the system gave results that more than confirmed all the hopes for it. At that stage of development the two data sets of cone and pore pressure were fed directly to a two pen TOA chart recorder without amplification. The only other apparatus were regulated voltage supplies working off a 12 volt car battery and a generator to supply power for the chart recorders. Figure C6.4 shows the machine in operation at Bafokeng. Figure C6.5 shows a section of a typical field chart result and Figure C6.6 shows this in the form of cone pressures against depth and also pore pressure against depth. The implications of the results justify detailed discussion of the chart output. The two pens of the chart recorder are offset so that they may pass one another, which means that the cone and piezometer traces are offset. After each metre of penetration a further rod has to be added and this takes about 45 - 60 seconds before penetration can restart. The chart was normally switched off for this period and restarted simultaneously with restarting penetration. The cone and pore pressure readings, and changes to them, continued in the rod change period, and it was observed that neither necessarily returned to zero. CHART RECORDERS, AMPLIFIERS AND CONTROL BOX. Fig. C ·6·4 (a) CPT RIG AT GYPSUM DAM - SHOWING OPTICAL ENCODER ON TRIPOD. Fig.C'6'4(b) CPT RIG AT PLATINUM TAILINGS DAM - BAFOKENG. Fia. C·g·4(c) , 10 II IZ 13 "" ". ', .. ' " 'i'} ( ~ ' I ",. . .... ,'" . :, 1;,:::; l :'~ J" 1'" i I: j" ,:,ijii Ii:, ijl! : II: t; lU i:;; ,:" . .---" ! I; !i:;:; ~iJ ~~ i ~:~i! ,~! ! i .<' .. I • • . : , ,; II: I 11 I' I I ., t ~ ~, • : " • I . : ,. I I . , : : I ; : : . : I' , :;.;;,;. 'I : i ; I' : t , 11 Id I'll i; 'II !l11'111~ ,. , ' il:! d "-.,..:;· :l i!· ~;' I:; !' ,:I,' " I, ,, "! t!d;, · . :!i ,' !" I " ' 111, : ,I iI:;I; 100 I = u. \ -U;---- \ 200 Z PIEZOMETER CONE LOGS FOR BAFOKENG Fig. C· 6·5 1 Fig C'6'6 167 The following observations were made from the chart readings during penetration : • Both cone and pore pressure readings fluctuated very markedly presumably indicative of a multi layered tailings of different densities and permeabilities. • The peaks and troughs of the cone and pore pressure readings were precisely in phase and of opposite sense (noting the offset pen distance) • The magnitudes of the corresponding cone and pore pressures were inversely proportional to one another viz a small cone reading corresponded to a large pore pressure reading. • The chart distance for 1 m penetration was about 22 mm and since the penetration rate was intended to be 20 mm/sec ie 50 sec for 1 m, then either the chart speed was inaccurate or the actual penetration rate was slower than it should have been. The latter proved to be the case and was about 15 mm/sec, ie at the limit of the recommended standard rate of 20 ± 5 mm/sec. • The chart needed to be annotated at rod changes since it was possible to lose track of the depth, particularly if stopping and starting of the recorder was not well synchronised with the penetration. • It was possible to select high pore pressure layers, stop penetration and allow the chart to continue to record pore pressure dissipations. The cone readings returned to practically zero in these extended periods. Distinct differences in dissipation rates at different depths could readily be seen. • The pore pressure system with a 200 kPa working range was inadequate even at modest depths and the peak pressures often exceeded 200 kPa. • The pore pressure readings appeared to be considerably more sensitive than the cone readings in that fluctuations in readings were more rapid. Not all the observations were positive and the following aspects were considered to be sufficiently problematic to require improvement before the equipment could be used for routine investigations: • Both cone and pore pressure signals, particularly the former, required amplification if the equipment was to be sufficiently sensitive to measure small changes. • • • • 168 The cone load cell required a smaller cross sectional area to operate at a lower load range and the fairly marked problem of a shift in the unloaded zero reading had to be eliminated. The cable was vulnerable to damage and to electrical interference. The filter elements cracked after loading The time base for the chart led to difficulties in processing the results, ie the chart movement should be controlled by depth not time. On balance, however, the results were considered to be overwhelmingly successful; the information available on the layering and pore pressure conditions within the tailings dam was far superior to that obtainable by any other investigation method and, subject to the difficulties described, the operation was straightforward. There was no doubt that it was necessary to improve the mechanical and electrical aspects to a user friendly state. Processing the results was extremely laborious and it took ten times longer to produce a finished CPT log than to carry out the test in the field - no immediate solution was seen for this except, as a starting point, having the chart transport automatically controlled by the penetration depth. At this stage Eben Rust, a newly qualified civil engineer joined the author as part of the team involved with cone penetration testing. Over the years he became involved, under the author's direction, with the interpretation as well as the equipment development and field work. The piezometer cone equipment developments in the early years (1977 - 1982) are listed below and are also described in the author's papers given in Appendix I. • • • Mechanical components, particularly cone seals and cable gland improved to prevent leaks. Load cell range reduced by having central hole; material changed from stainless steel to tempered high tensile steel with higher allowable strain which gave a greatly improved load cell. Piezometer transducer changed from Kyowa model PS-2KA to PS-10KA, giving a pressure range up to 1000 kPa. 169 • Amplifiers separate from the chart recorders were built, together with superior voltage regulators for the cone load cell and pore pressure transducers. • The chart transport control was changed from time based to depth based. The recorder being used was designed to accept the chart transport rate either from its internal timer or from an external source. Numerous devices were tried such as a long brass rod mounted on the CPT rig with a spring loaded connector to the moving cross head of the rig measuring the changing voltage along the rod, and a similar 1 m long threaded brass rod with a moving on-off switch counting the threads. The final system utilized a commercially available linear optical encoder: literally, a black box which, through an internal rotating shutter system driven by an external pulley, produced 5 volt square wave pulses which were fed to the chart recorder - Figure C6.7. The external pulley was driven by a belt to a pulley fixed to the cross head of the rig and the black box was mounted on a survey tripod - Figure C6.4. The encoder pulley size could be changed so that the ratio of penetration depth to chart advance was controlled and was usually set at 10 mm/metre. This depth linked chart drive system has worked virtually perfectly since originally fitted. • The filter element was changed from a ceramic to a plastic material used in large diameter consolidometers (150 mm dia Rowe Cells). Both face and base filter position cones were manufactured. • • The piezocone cable was made less vulnerable to damage and interference by threading it through a transparent plastic tube. Adaptors for pushing and withdrawing the rods were made which improved the ease of cable handling at the rig. The equipment was used in practice to the extent that during 1980 a number of site investigations were undertaken not on an experimental basis but for routine investigations and some of these are are discussed in Part D. 170 However, some problems still existed with the equipment, the primary one being that data reproduction was very tedious. Some zero drift also continued to occur on the load cell and although this could be compensated in the data processing it was not entirely satisfactory. At this stage (1981) the author moved to the specialist geotechnical consultants Steffen, Robertson and Kirsten for whom many of the site investigations using the piezocone had been conducted. From then improvements were primarily to the electronics of the system which were made under the guidance of the author and designed and built by Rust. A digital, as opposed to the previous analog, data logger was built which included new amplifiers and a digital tape recorder with a transfer port to connect to an office based computer. An analog chart recorder was used in parallel to the data logger so that a visual output of the cone and pore pressure readings was available during the penetration testing. SIQnot Condition., Coble Glon PENETRATION RESEARCH TEST ) - · I'--~ I I W AN'I IEKERl< . K ~ E Y N ') ,\8 EDWARD S \ ~ SCALE : I: 20000 FIGURE NO - 0·4· I o .0 o 181 There are a number of different ways in which the steps could be defined and combined, but they necessarily use the same basic data, ie the field settlement records and the recent CPTU results. In addition to the evaluation of the compressibility and consolidation characteristics it was also possible to estimate shear strengths both undrained and effective from the CPTU results. Further comparisons were also possible of subsoil descriptions from the borehole records and from the CPTU using the Jones and Rust soils identification chart. D4.2 Back Analysis for Umgababa As stated earlier the first step in the process was to determine the appropriate field fl\, and Cv by analysis of the settlement records. This analysis is described in the following sub sections D4.2.1 to D4.2.6. D4.2.1 Drainage path length For the purpose of this analysis the section at km 13020 was selected since the 1979 design investigation provided most information at this section. The geological cross section that was drawn up from the CPTU results is shown in Figure D4.2. From this it could be seen that the subsoil consisted of the following profile measured from the top of fill : 0 7,0 m Predrilled 7,0 8,5 Silty sand 8,5 21,6 Clay 21,6 - 22,0 Sandy layer 22,0 - 23,1 Clay 23,1 - 24,0 Sand The 0,4 m sand layer toward the bottom of the profile (21,6 to 22,0 below top of fill) acted as drainage boundary. This was apparent from the profile between CPTU 2 and 6, which showed the layer to be continuous with a different type of clay below the sand layer. Further evidence of the fact that this sand layer was a drainage boundary is the dissipation test carried out at 17,0 m in CPTU 2. This test showed that there was no km 13,020 CPTU 14 CPTU6 CPTU 9 .CPTU 2 CPTU I I' , ___ °1 .. . s.::: ~I~..!!!'.!:... ____ L-_______ ,. .. ' . .... , ... .. ' . .. ·· .. 1· ··· .. ' .. , '. - ' . . " " ;, ,',.,' "' :.' - '~ .:':':",:' ...: , ":,,,:.-, ..~::-: :~~.~~ ,,:,,:" : ' , ::.-:,:,:::.':,' >: :' :'-: :'::' ~ ', ' :.". ::.:,: : " " 5 -:" . ~:.:..~. -~ . ' ~'.:-: ~"±'-tt. .....,-.=""':"'. :--. ~.- -- 10 ~ CLAY ~ ' . ' ... ~, :- /0 '/ / / / ~ ---~ --.:.:..:...: '~-::::. -:::.: ,. ' .. '. '. .. ~ ~ ~.-- - -::::::::=:=::~::;:=::;::4:;;~~3::::!:::::::::;:;::'::::'::.:-::::.= _~~-:"'t..=-..s;-. t . C • • / ,, ' /. . / ~ /~ . ~ 15 - . . I", • ' . "'-4'~~ '. .. ;' • '- . . ... . ' . ., :, .•. .'" ". ' .. ' . - CLAY '. ..' , . ' '',,' 1I{"''=:\1\'~ '''' , . ' ' . ' . . .... " . , _ " .. . .. . " , " _ . . ,,,,~.l,'J'~ .. , "" "" ' " , " ' . •• SAND " '. '¥ - w' ''l , 20 25 ,~·1;"" )'\~g;. 1;;,,,-,,,, y.. . " . . , . .. . '" " • -: ," ,,' .. .. .. , .'.. , .. - , ' ''. ' : , , " , . .. ,", J\t.,IVg''''''''''t!lj.i '·"'''' ',".''''l=!!''J;''~/H.,("" c . ;; " ' .. '.' • . ' : '. '.: .' . . .. , " , . . , . ,:,,' . ' _. . . ' ". ' " ' , _ .. ' . ..'.' ' ,' ~ ,,,,e""'""~"~%"'»ll;'"'i;\': . -.~ ·'''-''·''.·A,''~ > ,""'" ~"''''h__ .. . - '''W<"'oc'' • • "-",, C ; ,~ '1,,' <' -"''''' '''~' "." .", .. ''' . ~"~,, . _" "''''''7I~ II 1~' ""C", . '·~<"""~"":rt ·:;%'::', r "'-;!;~a"""b~ ""'" ~'"'!=-""""'!,"{'''"~'''' """"·!l"!.;""'~'k'''",'t."'\".,.,, -" ~,c&o '''~'~' "i'~Ci-,t";"'~""-~''''':t''''''4 ~""'l'W~i .. ':i;iJ"~~'d"!(~¢}:{,i?;; '1_:::;i/~~~;~~~{~!~;i~W~~,,~!~,1~~~)~,~i~f!f,'~I"'I'ii~~~}~~~I~I'JrI~}~II,~~~~)kI~~I·~I~§\\~&.~~~~~~~\.~~~TtII I'{~~},~~/~~\;-I~~,\t$1f., ~_~jj.i{;1~,'?;;fl\<~&~1~~~i~~~i~~I 'IJalll==/'~j,z'~~;~~;~t1r~i~~lj~: .. - "".'- _ .. ' ''W .. , ••• ",,'" e"w":it" ',_ .. " "~'~' .> _ ".. "''';'' "_".. _ ,.. •. ", . """ , _,. "" .. ,_ " . ,,~ . .. " .. ~ __ ., .. '" " ... --, '-., ",~"".,,.; I'~'''~''~''''''';-~"~"i"~Q """~."a",,,,, -". -..... ," . '''~ >, .. ",,,,,, ~",,>ff, ~.11 • • ~" r,.;", .. , <'., "'",1" J<'~;'"",,,,,E>i",,..,,~, '~''''''k'''' . "· ... ·.t;1Jk·,,,,,,~'" N.- ..... ' , - '''''"'=f<:","" ,,, ...... 5".t.!lJ'¥; '"' ~ ,"''''~ !!; ~%i,' f'~7:"'~~ _ . '--~-:;'1"'~\i' ';:-%'li;~I!,,,;~~~~--;,U~l\. ::r. ,..:.,;;:, ! .... " .. NATIONAl.. .. OU l l IiIIAIiONAl( ROfTf GEOLOGICAL SECTION Umgababa km. · 13,020 23,24 CONE PENETRATION TEST PIEZOMETER (CPTU) RESEARCH NGL SCALE : HOR. I : 400 VERT,I : 2.00 FIGURE ~. 0.4.2 183 excess pore pressure in this sand layer. Therefore it could be concluded that double drainage existed, that the drainage path length was 6,55 m ie half of the distance between the top sand and the lower sand layer, 8,5 m to 21,6 m, and that the total thickness of compressible clay was 14,2 m. D4.2.2 Loading The final design height of the embankment was 5,6 m above N.G.L. The settlement vs time data indicated that 1100 mm of the settlement took place before the final grading of the embankment. Therefore the total amount of fill at 13020 was at least 5,6 m plus 1,1 m ie 6,7 m. The response of the pore pressure in the clay due to loading was governed by the following equation : D4.1 If it is assumed that a01 • a03 = 0, ie no shear stresses are generated, then au = D4.2 From the Rowe Cell tests the B parameter was established to be unity. Therefore: au = D4.3 This meant that the initial pore pressure at instantaneous loading was equal to the stress placed on the clay by the embankment. This stress was calculated as follows: aO = Ynat X h D4.4 = 22 x 6,7 = 147 kPa where Ynat = the natural (unit) density of the fill material estimated to be 2200 kg/m3 h = height of the placed fill (6,7 m) 184 Because of the width of the embankment no significant attenuation of stress with depth occurred, therefore at to (30 June 1980) : au = 147 kPa D4.2.3 Settlement record Settlement records for the Umgababa embankment were taken during and after construction. Figure D4.3 shows the settlement up to February 1990 when it was 2,690 m. D4.2.4 Rate of settlement A series of settlement beacons were placed on the final road surface at the inner and outer edge of each lane. The level of these beacons were regularly measured by relating the levels to existing fixed benchmarks. Typical results of these surveys done in 1988 and 1989 were plotted to show the total settlement - Figure D4.4 as well as the rate of settlement in mm per month during this period, Figure D4.5. The latter showed the maximum rate of settlement to be 7 to 8 mm per month in 1988 - 1989. D4.2.5 Degree of consolidation Typical ambient pore pressures obtained from the dissipation tests are shown in Figure D4.6. The ambient pore pressure at 15,5 m was measured as 205,8 kPa with the water table at 3,3 m. Therefore the excess pore pressure, ue' is given by the following: = 205,8 - (15,5 - 3,3) 10 = 83,8 kPa This test was carried out at 15,5 m which is 7,0 m into the clay layer. The normalized depth factor Z = (z/H) Therefore Z = (7,0/6,55) = 1,07 where z = depth into the clay layer and H = drainage path length 4 \ 40 35 :2 C 30 0 E --E 25 .s W I- 20 ~ I- 15 ill ~ W 10 -l ~ 5 en 0 16 E ~2.5 / § 2 -~ " -I~_=El/l/l++=99~~I_-l_~ _ _ II_-1 __ 1 _ ~ 0.5- --- t-- GEmEWEN'lM,l( ~ 0 4 6 8 10 12 14 0 2 DAYS (Thousands) UMGABABA II • ACTUAL SETT. BEST FIT SETT. SETT. RATE II Consol idat Ion Model 10 ~ r-', ,~ 0 \\ ~' I/"-F=7 ---/\ "'--V ~., -10 ~\ 't\\ I I / E " ....... ..I f .s ·20 \ /. ... ~. \ \... --.// \/...-""" ! I- ·30 \\ /1 z w / ~ UJ ·40 \ ~ / \\ .. I -l ~ .... --- ", UJ ·50 \\ \ / en / ·60 \ V \ / -70 V I V ·80 12.8 12.85 12.9 12.95 13 13.05 13.1 13.15 13.2 13.25 13.3 DISTANCE (km) UMGABABA SETTLEMENT (Thousands) II ( 1988-1989) 1- 1{7J88 - 21/10/08 - . 9/3/89 12/6/89 N'Be - ' Outer Edge 1 0~--~---'r---.----,---,,---~--~---,,-__ ,-__ ~ OESTFIT H TE 3CV6I198 • . 2-1---+--- 1----I-----I- ----I----I-----+----+ __ --Ir---i 12.8 12.85 12.9 12.95 13 13.05 13.1 13.15 13.2 13.25 13.3 UMGABABA SETTLEMENT RATE (1988 - 1989) NBC ' - Outer Edge DISTANCE (km) (Thousands) 11 '-- 21/10/00 ._-- 9/3/09 - 12/6/89 II 1 I I Fig. 0·4·4 Fig . 0·4·5 E I I- a... W 0 25 22.5 20 17.5 15 12.5 10 7.5 ./ /" ./ ......... ./ l--- / ./' ./ // -"" j ./ ./ \cLAYL YER o1----l----~--~--~~--4----4----li7~5--~2io~0--~2±2~5---j250 o 25 50 75 100 125 150 PORE PRESSURE Uo (kPa) UMGA8ABA - PROBE 9 Ambient Pore Pressure \1--- AMBIENT PORE PRESS. HYDROSTATIC PORE P. n 16--------------~--.---.---_.----._--._--_r--1T 4000 141----l----~---~--~----~--_+----~,-.,-/7 .. Lf----+---it 3500 3000 2500 2000 1500 1000 500 o 2 4 6 8 10 12 14 16 18 0 20 DEPTH (m) Fig. 0·4 ·6 E .s. I- Z w ::.:! UJ -l 1= w en UMGABABA - PROBE 6 CPTU . Cumulative Settlement II--qc -- SETTLEMENT II FiO· 0·4·9 0~~t==+~~~~~~~~==~~~~ __ 1~ ____ ~0 o 2 4 6 8 10 12 UMZIMBAZI - PROBE 16 CPTU Cumulative Settlement II--qc DEPTH (m) .. _-._. SE rTlEMENT II Fig. 0'4.10 187 The consolidation ratio at 15,5 m, or at z = 7,0 m, ie Z = 1,07 can be calculated as follows: = (147 - 83,8)/147 = 0,43 From the conventional Terzaghi consolidation chart Uz vs Tv and Z, Figure D4.7 it can be seen that: where Z = 1,07 and Uz = 0,43 then Tv = 0,33 Degree of Consolidotion( Uz) Figure 04.7: Degree of consolidation v depth factor From the usual equation relating U% and Tv or from Figure D4.8 it followed that the average degree of consolidation 0% was only 64,2% at the time of measurement, June 1989, some ten years after construction of the embankment. 0 .10 1::1 20 c .9 30 "0 1\ \ ~ 40 .0 on c 50 8 ., 60 '" 0 C 70 ., ~ 80 ., 0.. 90 "-~ ~ '--..... l---. t-- i--- --0..:::: - :---100 o 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 1.0 1.1 1.2 Time Foctor • Tv Figure D4.8 : Degree of consolidation v time factor 188 D4.2.6 Consolidation model A consolidation model was fitted to the actual records of settlement, rate of settlement and degree of consolidation. Classical consolidation theory was used to establish the governing consolidation parameters. The physical characteristics of the Umgababa embankment were as follows: Drainage path length Compressible layer Stress increase = = = 6,55 m 14,6 m 147 kPa It was estimated, on the basis of the cone penetration testing, that 96,5% of the measured settlement took place in the clay and the remainder in the sand. It was assumed, in accordance with Terzaghi's theory, that the consolidation parameters Cv and ~, as well as the drainage path length, remained constant with increasing stress and strain. A spreadsheet approach was adopted to enable multiple iterations of combinations of Cv and ~ to be carried out. The solution is given in Figure D4.3 showing the best fit for the settlement and rate of settlement. The best fit solution gave the following consolidation parameters : Cv = 1,5 m2/year ~ = 1,8 m2/MN The fit was clearly satisfactory despite the limitations imposed by the conventional simplifying assumptions of constant ~ and cv• The model predicted that it would take 24 years for 90% consolidation to take place and that the total settlement would be 3,86 m. It followed that a further 1 m of settlement would take place over the next 14 years. These predictions for the total settlement and the time for 90% consolidation were very close to the equivalent predictions made in the comprehensive back analysis conducted by van Niekerk Kleyn and Edwards (1985). 189 D4.3 Settlement and Time Settlement Predictions from CPfU at Umgababa The second step in the back analysis was to correlate the results of the CPTU with the back analysed fi\. and Cv given in the previous section. CPTU data was required for the subsoil both under and outside the embankment. The former, as described in the previous section, was in order to estimate the excess pore pressures and hence degree of consolidation, and the latter to assess by interpolation the virgin conditions which were originally under the embankment. D4.3.1 Compressibility correlation For this purpose a total of eleven CPTU's were carried out at the Umgababa site, four of which were considered to be outside the influence of the embankment. The results of these as average qc in the clay layer are : Probe 2 , qc = 0,447 MPa Probe 6 , qc = 0,431 Probe 7 , qc = 0,465 Probe 14, qc = 0,453 Average qc = 0,449 MPa A typical CPTU result (Probe 6) is given as Figure D4.9. The qc values increase with depth, but provided the layer thickness and depths and the rate of increase in qc with depth are similar for all the probes, then it is valid to use average values for the full depth, because the relatively large width of loaded area compared with the depth imposes stresses in the clay layer which are practically constant with depth. 190 The back analysed fl\, for Umgababa was 1,8 m2 jMN then since: _1_ = am'Ie D4.5 fl\, and fl\, = 1,8 m 2jMN and 'Ie = 0,449 therefore am = 1,24 As a check, this value of am was used to re-estimate the settlement for Probe 6 using the actual cone pressure data for each digitally stored reading ie at 20 mm depth intervals. The calculated settlement was 3,55 m compared with the measured (plus future estimated) of 3,86 m which was satisfactory noting that the Probe 6 average qc (0,431 MPa) was lower than the average qc (0,449 MPa). D4.3.2 Consolidation correlation The average CPTU measured tso was 37 minutes and the back analysed Cv was 1,5 m2 jyear. Therefore using the form of equation given below: Cv = cone time factor D4.6 tso Cone time factor = 1,5 x 37 = 55,5 min 2 jyear This compared very closely with the Jones and van Zyl, (1981) value of 50 which was to be expected since the initial value (50) was derived from data from the original site investigation at this site, and from similar sites, and from cone dissipation tests subsequently carried out in 1980 - 1981. D4.4 Application of Umgababa Derived Parameters to Umzimbazi and Umhlangane The third step (see Introduction D4.1) was the application of the constrained modulus coefficient, am' (1,24), and the cone time factor (55) obtained from the Umgababa analysis to the other two embankments viz Urnzimbazi and Umhlangane. 191 The fourth step was the comparison of the resulting predicted settlements and times with those measured at the two embankments. D4.4.1 Umzimbazi (a) Settlement The CPTU showed that the average qc value for the probes outside the embankment was 0,316 MPa. M = 1/fi\, = = 1,24 x 0,316 2,55 m2/MN The original thickness of the clay layer, H, measured by the CPTU (and compared with the original investigation) was 6,95 m. Therefore the estimated settlement was: aH = aoxfi\,xH D4.7 where a 0 is the stress due to the embankment at the centre of the clay layer : this is approximately 100 kPa since the fill density is 2200 kg/m2 and the fill height was 4,6 m. Therefore: aH = 100 x 2,55 x 6,95 = 1,77 m (Figure D4.10) The measured settlement at this section in 1985 was 1,75 m. This was obtained primarily from the settlement records, but also from the CPTU carried out through the centre of the embankment which showed that the present clay thickness is 5,20 m ie 1,75 m less than the clay layer thickness outside the embankment. 192 (b) Consolidation The CPTU gave an average t50 of 22,8 minutes, hence the coefficient of consolidation derived from this, using the Umgababa derived cone time factor of 55 is : jj 2/ c =-m year v 22,8 c = 244 m2/year v ' Back analysis of the settlement records gave an average or field C v of about 4 m 2/year. Figure D4.11 shows the actual settlement data and a best fit single value Cv and ffiy plot. D4.4.2 Umhlangane (a) Settlement The CPTU showed an average qc of 0,563 MPa hence : 1 M = _ = 1,24 x 0,563 mv ." IC o ~ CJ> CJ>f'T1 ",-f -f-f -fr rf'T1 f'T13: ;:f'T1 f'T1Z Z-f .;-f l> c Z ;:0 N::u ;:~ CDf'T1 (.:jO ." 1.: I I \-1 ·---.I'----I---~---I II -----=t i I- 1.6~ ~.-~-~ ~·----r-.-.I~-~-l. I BEST tiT SETTLEMENt __ 1.4~ I 0-,--L--J.---[---- . E ----~ 1.: _ 1--\--·--[--~··]=~r--r--·--·t---rc: G 0.8 ~ r\rl- -r--~·-----1-.L1/-1/1..,..-99--.~J-----+- (f) 0.6 - I 1-\.\ i'-r--r-'l--'--I--- ----.:\\ .. J~----I---·T--T--l----,---+ I- ---f------ 1_:.., ___ . __ . __ ._._ .. , ... ___ ._. _______ .. _._ ... 1 .. __ ....... ____ . ___ . ____ .. I _____ . _____ L ____ ._ .. --- I """"', . ,--'---_ .. _--.. ---. I I I I : o I I i I I _nr-.. --.. ·-_-.·m .. j--- I I i I 0 ! CPTU SETTLEMENT I-t- 20 18 --£ 16 ~ c 0 14 ~ E 12 5 W I- 10 « a: 8 I- Z W 6 2 w -.J 4 ~ W (f) 2 o 500 1000 1500 2000 2500 3000 3500 4000 4500 5000 DAYS - ACTUAL SEn. - BEST FIT SEn ........... SEn. RATE - CPTU SEn 194 The thickness of the clay layer was 12,0 m, the embankment was 7,8 m high with a density of 2200 kg/m3, therefore the settlement estimated from the CPTU was: aH = aa x m., x H = 171 x 1,43 x 12 mm = 2,94 m The measured settlement was (1990) 2,79 m and the rate of settlement was about 1,0 to 1,5 mm per month. The CPTU showed there was a small excess pore pressure, and therefore it was reasonable to assume that part of the current settlement was consolidation and part secondary. (b) Consolidation The CPTU gave an average t50 of 30 minutes, hence the CPTU derived, Back analysis of the settlement records showed that reasonable data fits could be obtained with cy in the range of about 2 - 6 m2/year depending on assumption made regarding the present small excess pore pressure and the contribution of secondary settlement to the total. D4.5 Summary of Results The settlement and consolidation prediction data is summarized in Table D4.1 together with typical laboratory test data from the original site investigations. 195 Table D4.1 : Measured and predicted settlement data Test Method Umgababa Urnzimbazi UmWangane Coefficient of compressibility, 11\, m2/MN Lab 50 mm 1,67 0,82 . 0,53 Lab Rowe 112-224 kPa 0,93 1,06 0,47 Lab Rowe 224-392 kPa 1,53 0,92 0,39 CPTU predicted 1,80 2,55 1,43 Performance measured 1,80 2,36 1,38 Settlement, m CPTU predicted 3,86 1,79 2,94 Performance measured 3,86 1,66 2,85 Coefficient of consolidation, cv' m 2/year Lab 50 mm 0,71 2,1 0,68 Lab Rowe 112-224 kPa 2,60 3,8 0,47 Lab Rowe 224-392 kPa 1,40 3,7 0,52 CPTU predicted 1,50 2,44 1,85 Performance measured 1,50 4 4-6 Table D4.1 shows that prediction of settlements at Urnzimbazi and Umhlangane using the am value (1,24) derived from Umgababa gave a remarkably close estimate of settlements. The laboratory tests average values of 11\ gave very poor settlement predictions. The higher laboratory 11\ values would, at Umhlangane, have given a reasonable prediction, but at Urnzimbazi a poor prediction. Table D4.1 also shows that the comparisons of CPTU predicted and measured coefficients of consolidation were better than the comparisons of laboratory and measured coefficients of consolidation. The conclusions from the 1989 - 1990 research project were unequivocal and are stated below: 196 (a) Settlement magnitude The data shows that the constrained modulus coefficient, (Xm' (1,24) backanalysed from the Umgababa data gives excellent settlement predictions for the two similar embankments at Umhlangane and Urnzimbazi. (b) Consolidation time The data shows that the cone time factor of 55 backanalysed from the Umgababa information gives satisfactory time predictions for the Umhlangane and Urnzimbazi embankments. (c) Undrained shear strength The research also confirmed that undrained shear strengths, su' can be derived from the cone pressures in the usual way and that the conventional values of Nk viz approximately 15, gave satisfactory values. ( d) Effective stress shear strength parameters From the piezocone data it was possible, in the apparently homogeneous clay layers of significant depth, to plot cone pressures against effective vertical stress and hence in a manner similar to that suggested by lanbu and Senneset (1974) to derive ",'. The data gave a value of 19° with a c' of zero. Whilst this is not very close to the laboratory triaxial test value of about 25° - which would not in any case be expected since the type of test is very different, as are the initial stress conditions - it does provide a means of deriving equivalent triaxial ",' values from the CPT for the typical estuarine clays at these sites: tan",' cp = 0,75 tan ",'te D4.8 where = CPT derived = triaxial compression measured More confidently, this approach can be adopted at any site to examine the variation of ",' with depth or position rather than to assess its equivalent triaxial compression value. 197 The most significant conclusion is that given in (a) which is that for the three embankments the constrained modulus coefficient for the soft clays is : (lm = 1,24 and that this value should be used to derive drained modulus E' (or M) values for the clays from CPT cone pressure values, qe' using : = D4.9 It must be emphasized, however, that this value of (lm is considerably lower than that given in the international literature which, for normally consolidated medium to low plasticity clays, and for qe less than 700 kPa is in the range 3,7 to 10 and for highly plastic clays 2,5 - 6. In other words using the literature values would have only predicted about one third to a half of the settlements which have actually occurred. Examination of the laboratory measured Il\ values given in Table D4.1 shows that they predicted only about half the total settlements which occurred at the Urnzimbazi and Umhlangane sites. Therefore although there is excellent agreement on the appropriate value of (lm for the three embankments within this project, it remained to be shown that the use of this value is valid for other embankments. The value is certainly appropriate for the prediction of total settlement for the three embankments (since it is obtained from back analysis of their performance), but the total measured settlements may have been due not only to primary consolidation, but also to other factors. Since at both Umzimbazi and Umhlangane partial stability failures took place during construction, then because of the high stress ratios the total settlements should be expected to include relatively large non consolidation components. Thus in cases of similar high stress ratios, (or whatever factors distinguish these embankments), the low value of (lm is appropriate for settlement predictions. It should be emphasized, however, that this value of (lm is for the prediction of the total settlement and includes components due to local yield, immediate, primary and secondary settlement. In other words it allows prediction of the medium to long term worst case scenario. 198 Correspondingly, however, for situations where consolidation settlement is dominant, presumably where lower stress ratios are applied, then the (lm value of 1,24 may be over conservative. The project concluded that an (lm value of 1,24 was appropriate for total settlement estimates for" highly stressed soft clays which typically occur in the alluvial deposits along the Natal coast and that a cone time factor of 50 gives a satisfactory conservative estimate of consolidation time. Further research was however, recommended at other embankments for which settlement records were available and a larger variety of subsoils were present, so that the applicability of the (lm value could be substantiated, or a range of appropriate (lm values obtained for different subsoils and stress conditions. This proposal was accepted by the Research and Development Advisory Committee of the South African Roads Board and a description of the resulting project is given in section D5. 199 D5 CPTU RESFARCH PROJECT, 1991 - 1992 D5.1 Introduction The objective of the project was to determine constrained modulus coefficients and a cone time factor with sufficient confidence so that they could be used in practice. The approach adopted was to increase the data base from the three embankments of the 1989 - 1990 project described in D4 so that a variety of subsoils and embankment situations would be included. The criterion for selection of the embankments was that adequate settlement records should be available. The methodology of the research was to carry out CPTU's at these selected embankments at positions which were not expected to be influenced by the settlement, but would yet be representative of the before construction conditions. This presupposed that original geotechnical investigation information would be available, and for those cases where the situation necessitated embankment monitoring this was almost inevitably so. Using the settlement records a back analysed coefficient of compressibility, ~ was calculated and this was correlated with the CPTU results to obtain an am for each embankment. In a similar manner a coefficient of consolidation, cy , was back analysed for each embankment from the time settlement records and this cy correlated with the dissipation times from the CPTU's. Where sufficient information existed from the design geotechnical investigation, in particular laboratory consolidation test data, then further analysis was possible ie "predictions" based on this data and on the CPTU could be compared. The expectation was that a range of am would be obtained and that the values would reflect the nature of the subsoil. In addition it was considered likely that the unusually low values of am deduced from the 1989 - 1990 research project could to some extent be due to the high stress ratios for those three embankments and the inclusion of secondary compressions in the measured settlements. It was also realized, however, that the selection process for the 1991 - 1992 embankments necessarily included only those embankments with adequate settlement records, and that there was a probability that these too would be potential problem sites and high stress ratios could apply. 200 A total of fifteen sites were selected from discussions with road authorities and consultants. In view of the time and budget constraints not all these could be utilized. A further selection process reduced the number of sites to eight at which it was believed the best records were available and which covered as wide a range of materials as possible. At some of these sites there was more than one embankment so that a total of sixteen embankment positions were included. At each of these CPTU's were carried out, so that a total of 35 CPTU's were performed. During previous investigations at these sites, and at the 1989-1990 research project sites, 60 CPTU's and 60 mechanical CPT's had already been carried out, with the result that data from total of 155 cone penetration tests was available. The project began in June 1991 and was scheduled for one year. Obtaining and sifting though the records to select the embankments for testing took much longer than anticipated as did the final assembly and evaluation of the data from the selected sites with the result that the project will not be completed until the end of 1992. The information and conclusions given in the sections D5.3 and D5.4 are therefore taken from an evaluation of as yet unpublished data. Brief descriptions of the sites are given in the following section D5.2. D5.2 Site Descriptions D5.2.1 Introduction Eight sites were selected and since some of these had either more than one embankment, or a long embankment over which conditions were significantly different, sixteen evaluation positions were available. The project sites are listed below: • Manzamyana Richards Bay one embankment • Umlalazi Natal N coast two separate embankments • Umhlatuze Natal N coast one embankment • Prospecton Durban area one embankment • Mzimkulu Natal S coast three positions at one embankment • Goukamma Cape E coast one embankment • Hartenbos Cape E coast three positions at one embankment • Bot River Cape S coast four positions at one embankment 201 Oescriptions of the sites are given in sub-sections OS.2.2 to OS.2.10. Typical piezometer cone penetration test logs are given in Figures OS.1 and OS.2. In addition, because construction at some of the embankments was divided into distinct stages resulting in significantly different stress conditions for each stage, it is considered valid to analyse these stages as separate cases. The total number of cases, including the three embankments for the 1989 - 1990 research project, is twenty five. OS.2.2 Manzamyana The project comprised an 8S0 m long viaduct to carry a 10 m wide harbour access road over railway lines and the Manzamyana canal. The approach embankment at the north end is only about 3 m high and has been monitored over a length of about 60 m adjacent to the piled viaduct. The recorded settlement over a five year period has been about 1,3 m which has necessitated remedial measures. The subsoils consist of about 2 m of silty sands deposited recently by hydraulic filling; over 10 - IS m of very soft silty clay (Lagoonal clay); over estuarine sands and silts; over firm to stiff silty clay (Estuarine clay); over silty sands; over stiff silty clay (Lower Lagoonal clay) to a depth of about 4S m - CPTU log Figure DS.l (a). Laboratory results from indicator, consolidometer and triaxial tests were available from the original, 1983, investigation. DS.2.3 Umlalazi The road project included two embankments relevant for the research project. One of these is the national road north approach embankment to the Urnlalazi river bridge and the second, about half a kilometre to the north, is across part of the flood plain of the Urnlalazi where a minor tributary enters. The embankments are each about 200 m long and up to 8 m high. The original extensive investigations (1979) resulted in the prediction of large settlements together with stability problems during construction. Vertical sand drains and sand blankets were incorporated to reduce these problems. Excellent monitoring results were available over a five year period from the start of construction until 1992 and these show that settlements of 1,3 m and 1,8 m occurred at 202 the first and second embankments. The subsoils at the first embankment comprise recent alluvium of sands, silty and clayey sands and clayey silts from 3 m to 16 m deep, the deepest being furthest from the present river course indicating that this had changed - CPTU log Figure DS.l (b). At the second embankment the subsoils are shallower, being up to 7 m deep, and comprise finer sediments of clayey silt and silty clay. Comprehensive laboratory test results were available from the original 1979 - 1980 geotechnical investigation. DS.2.4 Umhlatuze This site is unusual in that one section of the embankments for the national road across the wide Umhlatuze flood plain layover a deep peat deposit, whereas most of the embankments are over the more common silty sands and clayey silts. The extent of the peat had not been fully determined at the early investigations, (1979), but subsequent settlement measurements during the preloading stage of construction highlighted the problem along the section of embankment included in the research project. The relevant section of embankment is about SOO m long and up to 11 m high (including 3 m surcharge). Monitoring results were available, although many of the monitoring systems were destroyed when the embankment was partially washed away during a flood in 1987. The embankment has settled over 7 m (May 1992) and is continuing to settle at about 8 mm/month. The subsoil consists of up to 13 m of peat over silty sand and silty clay layers to a depth below ground level of about 40 m - CPTU log Figure DS.l (c). Comprehensive laboratory test results were available from a supplementary investigation carried out in 1987. DS.2.S Prospecton An interchange was required over the existing freeway south of the airport at Durban in the area of the Isipingo and Mbokodweni rivers flood plain. The national road is at existing ground level and the interchange structure approach embankments are about 8 m high. The area was developed from the flood plain over the past twenty five years to be an extensive industrial township by filling with about 3 m of Berea Red sand obtained from nearby hills. A history of foundation problems existed in the area due to settlements of the deep alluvial deposits. eaa IECJ:' ROAC INIC1 C?IU !l!i~"8Q! C2DJ (EEl' MANZJ ~. 19-19-1~1 S}lF. /l4AN~A/I4YANA "l"MT CONE R£SISWICE 0 0 2.5 ~) 7.5 10 ( I ............ I ) ..cr- 2 p- 3 1\ -=:::; 1\ I I ~ • \ L.=- 5 I -.l £. ===== S6 ?\ I ;;;7 I \ I ~ 8 I "-~ I 9 1 I 10 I '\ II I~ 12 I I ~ 13 tl I' ~ 15 - 100 ° 100 <)0 300 400 PORE PIlESStJIlE (ld'a) 0) MANZAMYANA peo.lECJ:·RQ'C INIC1 CPlV I$SE'8Q! w= MlAiUZI --- CONE RESlStANC!: 7.5 10 o r---~0-~-F2L5--+----r~-~ -i2.. I ~+----t==~~~ ~ I 2~---+--~4====-~----+----~ 3~'~~~~==~~==~ .~~r~ __ ~-+I __ ~I __ ~ )3 s~--4.---~I--~----+---~ '* g6 ~--'I~~'--+---~---+--~ ~7 ~--~.~\ ~--+--~---t---~ ~ \ I I 8~-r~,~~t--~1-----r---1 9~-~~7\-~1---+---r--~ 10 f----+l-"---'-----+~:-----+---+----I Ill---j:~~?:'--I----I-I I --'--" 121--I-~,,~~F::::=+====t=;:>:::-1 13 f-----j-~-d--_+I --_+_~~=1 I·r-+~"'F:=f=~~~ 15 -100 0 100 /00 '00 --- PORE PIlBSUlE (ld'a) c) MHLATUZE eRO.1ECl' BOAC ltilel CfIII ~s:EA~ CZIlI fREl' MI &2 S!IF' HI '1 AlT QAIf' 40_28_1991 CONE RESISt'\NCE (M~) 0 2.5 7.5 10 O. ~ ~ ! 2 3 • '1 I~ \ I'? 5 I------" S6 1\ L \ I 71 I ~7 \"...- I ~ 8 ,, \ I 9 11 I "', 10 It \ II I~ ~ 12 1< I ~ 13 It ), I. ~ ~ 15_100 ° 100 200 )00 ' 00 POll!; PIlESStJIlE (ld'a) b) UMLALAZI PRQ,1ffJ' BOAC INTel ePlU REC::ABCH cpnJ CRf)' PS09 s:JF: PBpsPp"r DATE' 11:/)1-1991 o 100 200 .00 --- PORE PIlESSl.I ~ ~ >::S: S D """"<: ~ ~ r ~ t:::= ~ -::;;::: ~- -I:;;?- '> ~ J 5_ 00 : ( o 100 /00 300 '00 --- PORE PRESSUlE (16'0) d) BOT RIVER TYPICAL PIEZOMETER CONE LOGS Fig. 0'5·2 21)7 The bridge was piled and shielded from lateral loading by stone columns through the alluvium. Settlements were monitored during construction and were only up to 0,15 m; nevertheless the embankment is considered to be relevant since it represents an example in the low stress range on relatively shallow alluvium with sandy clays of soft to firm consistency. Three of the positions monitored are considered to give a sufficient information to allow separate analyses of the settlement. However, because of the short time for settlement, the records are considered to be inadequate for any meaningful consolidation time deductions to be made. D5.2.9 Bot River The Bot River embankment is on a Cape Provincial Trunk Road and is the west approach embankment to the Bot River bridge. The embankment is about 0,5 km long with a constant height of about 6 m over the relevant section which is about 0,3 km long. The author, whilst at NITRR, assisted with the original site investigation in 1975, using the then new 100 kN imported standard Dutch CPT rig, with the standard mechanical friction sleeve cone and load measuring equipment. The investigation, which included boreholes, sampling and laboratory testing, the results of which are available, showed that severe stability and settlement problems were to be expected over the recent soft clay alluvium which is up to 18 m deep. The embankment was therefore constructed in two stages, viz an initial 4 m followed by a further 2 m, and in addition a 1 m high surcharge was placed close to the position of the proposed bridge abutment to minimize subsequent lateral loads and negative skin friction on the bridge piles. Extensive monitoring was conducted at four positions along the embankment and full records were available for the construction period from 1976 to 1981. Since there was a long delay between the first and second stages the information was sufficient to consider the embankment separately for these two stages. The settlements along the embankment have varied from about 0,4 m at the western end to about 1,9 m close to the bridge abutment. Only small and even settlements have taken place since construction and no remedial measures have been necessary. 208 05.3 Research Project 1991 - 1992 Site Investigation Ouring 1991 - 1992 each of the embankments described in the previous sub-sections was investigated using the latest piezometer cone penetration test equipment. This, as described in section C6, included a lap top computer in the field which enabled immediate production of cone and pore pressure logs and the creation of data flles which has facilitated subsequent data processing. A total of 35 CPTU's were carried out at the eight project embankment sites and typical logs for each site are given in Figures 05.1 and 05.2. A total of 75 pore pressure dissipation tests was carried out. In addition to the information from this field investigation, both field and laboratory test results have generally been available from the geotechnical investigations carried out during the original road design projects some of which were more than fifteen years ago. It was a prerequisite for the selection of embankments for this project that settlement records should be available. In some cases excellent records were kept and two examples are given in Figures 05.3 and 05.4. In most cases the records have not been so comprehensive although they are adequate for the present purpose. 05.4 Methods of Analysis of Results The primary purpose of the project was to determine appropriate constrained modulus coefficient, (tm' values for the South African recent alluvial deposits by comparing measured settlements with CPT data. The secondary purpose was to obtain data on the consolidation time characteristics by carrying out pore pressure dissipation tests. From these t50 times coefficients of consolidation could be estimated and compared with coefficients of consolidation calculated from back analysis of the measured rates of settlement. Derivation of am In addition to comparing the piezometer cone penetration test predictions with the measured embankment performance it was also possible to compare these CPTU 209 predictions with predictions of settlement and times for settlement made on the basis of the original site investigation laboratory test data. As discussed in D5.1, the method of analysis of the data was essentially a comparison of the new CPTU results with the back analysed coefficients of compressibility, 11\., and coefficients of consolidation, cv' for each embankment in order to obtain (lm and cone time factors. The data set consisted of twenty two cases from the 1991 -1992 research project, together with the three sites from the 1989 - 1990 research project. The subsoils are all recent alluvial deposits, varying from silty sands through sandy silts, clayey silts and silty clays to peats. It was anticipated that because of the wide material variations, and the cone pressure variations within any material type, then the (lm values could also have a wide range. If this was so, then it was hoped that the variations within this range could be rationalised using the materials information from the original site investigations viz natural moisture contents, void ratios, plasticity data, gradings, compressibility and undrained shear strength, together with initial and final stress ratios and the measured cone pressures. Section B5 discussed settlement analysis for embankments and it was noted that the total settlement comprises components due to local yield, immediate settlement, primary consolidation and secondary compression. Although the international literature is not explicit on precisely which settlement parameter may be derived from CPT cone pressures, the consensus is that it should be the coefficient of compressibility, 11\.. The implication of this is that it is only the immediate and primary consolidation components of settlement which are represented by the cone pressure derivation and that the remaining two components must be assessed by other means. It follows that back analysis of the measured embankment settlements must separate out those portions of the settlement due to secondary compression and local yield. In most cases it has been possible to separate any secondary compression because adequate time settlement data is available to allow modelling using Asaoka's method. This defines the end of primary consolidation, hence settlement in excess of this may be assumed to be due to secondary compression. From the settlement data available it is not possible to separate the non secondary compression into primary, immediate and local yield (where this is relevant). These can only be subdivided as described in section B5 by making various assumptions regarding the stress conditions, the relative values of the drained and undrained moduli and Poisson's ratio. o o z ~ ::0 C o -i (5 Z l> Z o en fTI .... -i f;; 3: ", Z -i ::0 ~ i 8 7 6 -.... ~ 5 ~ ;::: 4 ~ "" ~ 3 ...J ~ 2 l( I o ,,0 .,0 ./ /0 0/0 /0 .; b 200 LEGEND Fi II Elevation. BOT RIVER STATION 8 + 830 Plate nOo4R ,./0..,..,0-0_0- _0-,0 --0-,0- 400 600 BOO No. OF DAYS SINCE 9/05/76 Plate Settlement. _._.- 1000 2,5 21) -. ~ '- to-- 1,5 ~ ~ ~ 1,0 ~ ~ o~ ~ r"~ ,0 1200 10 g E II 7 .... x II ----~ ... 5 x 4 oJ oJ 3 ... 2 0 10 II • oJ (/) ,. 7 I 6 oJ ... 5 > .. J 4 0: 3 ... ~ 2 ~ 0 0- 0,!I I .... z .. 2 I , ... .J .... .... '" VI I", FILL CONSTRUCTION ------------------ - ---- ------------------------- ----- ---------- I I PIEZOMETERS ~ .--=--~"'-------------- .--- ---------:--:..-- -- ;--- ... -# ..... ~~_/ ./~- ------.:::-=-::------- -- r" /."" -- __ .:----- _ _____ _ /::.~1';.~;~~~7--,d-!/ ~ ~----=- .- --------~-----.-- -- '--=- - - ~_~~ ____ ~~~~.:::;~<----- -'----;;~_=_ :-_=_~~=-~~--::. -- ~ __ ... --:---:----~-----;-h--....~----:a--..:..~ ,----------- ,,;-/ ... --'~- - ... - - .. - "- "' - '. ........----_...-- - - :------ ------------------------ '~ .y-----~./- ~~.~.--=====--~==~~~~:;;;;.;~~~.~;~~~=~=~===~~=====~==~==~~~.~~.~. ~.;~.~ .. ~~~. .. ~ .. ~ . --------~ ..,;;:.,_", c- ---~ . . ......... .... ..... . ............. - . . ...... r;::;;......... • •• -- ___ I ~ r .0 .................. -.0 -.' J' • / '\. ...... • • •• :..:...,. =~'':': ~::::: :.;;: ~ ___ ','. __ 11 _ .. ~ ••••••• • ..r-1I. .-._.__ .r.... ~ ... .................. -.............................................................. _ ...... -................................................. .. .... _ ...................... P .................................................... P .............. . -- ~ .... UA-rr"" , _ -<.... ~ '-. ../. -.-/ _ -.-------..:.::=~-.---.-.-.-.- . -.-.: . - ~ . -.- .. '-Y·'-.-:..;- .~·-·"""", . ~~-.- ._._ ~ ,/ '- • ....,,' y.. ........ / .. -. -.- _ • .r--._. - "- ' _JI ........ • -.._~~_ '-'~_'_ ._._._ ._._._ . ____ _ ~/., -:J... ' / -Il_,. ___ .. _ .. -==--~_ ._II ___ ._ .. _I-...,.._ •• _~._ .. ,'-_. __ '='.:::::";.- .. -IC-,,-.--tt--«_II_Jr_. ___ ._. ____ .. _ .. _,,_,,_._,._.-_. ___ .-.-_-II-.. -.-.-tt-II - "'--'_ 1I _ 1I_'_"_"_"_"_I/(_,,_' ..... <'J" ...-. -._.- - .- - .- - - - .- .-.--'-.. .- .- SETTLEMENT RODS '-'- "- - - - SR 5 ' ....... ~~ ---- SR 7 -- "... ...... ~- ........................ ----=-~- .. ;:::-.-:;:: .... '-...... ..... ;::;- ... :. --:::-...... """"' -- ...... :. ..... -.:--~ .'.< ... '-"~ -- SR 8 .......... SR 9 ------ SR 12 --- SR 13 --- SR 14 ........ -... . .•... ~.:::.:::.:::- .... ..•. -......: ....... ~ .' ••••• ::.;;;..~::;:-" _, _ _ _-;:. - --; BUR rED ...... ::~ -- - -------____ ~==_=_--.,-..... --+---- 3 IS 2,5 t: u ~ .. G\..ob t ... D .. 8 2 • ~ ®~ ",e0 , 0 • ;: « 0,6 cz: V> • V> .... • cz: 0,4 l- V> • • 0,2 I • • \ .. : • •• • • •• • • • • • • • • • • 0 • 0 I 2 3 4 am e) Stress ratio, ~ H/q • v constrained c modulus coefficient, d m Fla. D· 5·6 222 DS.5.2 Coefficients of Consolidation, Cv As indicated at the beginning of this section the secondary purpose of the 1991 - 1992 research project was the derivation of consolidation characteristics from piezometer cone penetration testing. The field technique adopted was that which is now accepted internationally and first advocated by the author in the literature (Jone and van Zyl, 1981). Cone penetration is stopped at selected depths and the time for half dissipation, t5O, measured. Half dissipation tso is defined as the time at which the initial excess pore pressure, uo'( at time to) has dissipated to Uso half way towards the final pore pressure, U 1OO• At all the embankments in this project the water level was close to ground level hence u100 is hydroststatic and is simply calculated from the depth at which the dissipation test is conducted. The value of U50 can therefore be readily determined as soon as Uo is registered on ceasing penetration and the subsequent time taken to reach pressure Uso is measured. The author's method of estimating the coefficient of consolidation, Cy, was based on the empirical equation for the South African alluvial deposits viz : = DS.2 where c;. is m21yr and t50 is in minutes Since the alluvial deposits have coefficients of consolidation in the range of say 1 to 100 m 21yr the half dissipation times vary from about an hour to one minute. The c;. values obtained from the CPTU's are then compared to those derived from back analyses of the embankment performance. The latter are calculated from the Asaoka plots of settlement. This is an extremely useful technique and has the major advantage that c;.'s can be estimated before the end of consolidation. An explanation of the application of Asaoka's method is given in Appendix III and examples of the use of this method are shown in Figure DS.S. In these the end of primary and beginning of the subsequent secondary consolidations can be readily distinguished. The number of points and the fit of a straight line through them give a good indication of the reliability of the definition of c;.. This method has the further advantage that the difficulty of analysing settlement records during construction by artificial curve fitting is largely eliminated. UMZIMBAZI UMGABABA Asooko Asooka /600 L 2000 / 1600 .#" 1600 y;/' 1600 ~ 1400 ~/ : 1+00 // . /200 /7 ! IZOO ~ : ~/ :: 1000 ~/ " .. \.6'000 / - / 800 ~ /' 800 /- / 600 / /' 600 / / 400 /' 400 / ZOO toO /" olL" 0 200 .00 600 800 1000 ,icc 1400 16<» 1m 0 200 400 600 800 1000 "''''' ''''''' 1600 ,800 2000 0 6 (i-lJlIP/1t 6(1-11"..." - MZIMKULU PROSPECTON Ascoko Asooka zooo / /800 ~ /" 1600 / 1400 _/ ,- 1200 ~ ~ ~ 1000 ~/ ~ -~ ~. /' ~ 800 , /'" / '" 600 / +00 V / ../ /00 / , zoo lL 0 0 0 200 400 6 00 800 1000 12(;0 , .. 00 1600 Itm 2000 0 100 200 300 400 !KXJ 600 ((i-I) "'". i (; -/J ",,,, - . - GOUKAMMA UMLALAZI NO. 2 . Area C zooo .,., /" ",--;;/ 1600 -' .// 1600 ""0 ~7 ~ff ' 10 10 10 > qc > 5 20 5 > qc > 2,5 20 2,5 > qc > 1 15 1 > qc 35 It will be seen that 50% of the readings at these si~es were below 2,5 MPa and it can be seen from Figure 4 that the Begemann chart is difficult to use at suc~ low cone pressure. This does not of course imply any deficiency in the basis for the chart, but merely serves to demonstrate that the cone pressures at the sites tested are generally extremely low. A further disadvantage of the direct use of Begemann's results for the particular purpose des cribed here - i.e. preliminary estimation of settle ment under embankments - is that in soil testing for roads particle size distribution below the No.200 B.S. sieve size is not usually carried out. Hence the % < 16 \I size is not a very convenient parameter. The present investigation therefore examin:d the possibility of correlating other parameters aga1nst friction ratio. The results are given in Figures 5,6 and 7 in which for 140 samples, taken from three different sites, the friction ratio is plotted against % < No. 200 B.S. sieve; % < 20 \I and Plasticity Index respectively. w > w 100 o 130- & d Vi 60 '-- U1 aJ o o N o Z o d (.!) z Vi 40 _ d en !>!> x % <20~ -0,44, ° /.. " '),89 N'U':l ~/. ----REGRESSiON liNE FOR TESTS USING IMPROVED fOUIPMENT , /. ~/. , .I.: /. 20 MOO I. '0,537 x 0/0< 2 O!,- - J,00; , , 0,85 II' 4 5 ----- COM8INF.0 LlNf 40 MOO I. C.55 x 'i'o <~0l'- - 2,36, ' .' 0 , 88 N.'183 60 eo % <:: '20fJ- FIGURE 9. RELATIONSHIP , BETWEEN MODIFIED PLASTICITY INDEX AND PERCENTAGE SOIL < 20 \I 174 A statistical analysis gave the following least squares linear regression equation for the total of 183 samples tested. It is of course valid to use the results of the laboratory tests on both the old and new samples since this comparison is not concerned with sounding measurements. Modified Ip = 0,55 (% < 20 \I) - 2,36 ------- (7) with . r = 0,88 The regression lines for the earlier and latest sites separately and combined are shown on Figure 9. It is not suggested that this is in any sense a valid gene ral relationship for all soils but only that it holds for recent alluvial deposits in local conditions and that for the purpose described here it is useful. The Begemann line, shown on Figure 6 as a plot of friction ratio against \ < 20 \I, can then be trans formed to a plot of friction ratio against Ip (Figure 7) by simply converting the \ < 20 \I in Figure 6, for any particular friction ratio, to an Ip by using equa tion 7 given above. It can be seen that the agreement between the latest line and the Begemann line in Figure 7 is very close in the linear range. There is a little diver gence at the low friction ratios and rather more divergence at high friction ratios. However, bearing in mind the purpose of friction ratio measurement, which is to provide a reasonable description of the material, these divergencies are not of great signi ficance. If the friction ratio is outside the linear range then the description is well defined either as clean sand if F.R. < 2, or as medium to high plasti city silty clay if F.R. > 6. For convenience a chart, Figure 10, showing fric tion ratios against material type subdivided into zones, is given for use with local alluvial deposits. ::t o (\J V o!! 100r-------------------------__ -, CLAY <'0 60 SILTY CLAY SANOY CLAY 40 CLAYEY 20 0 J 4 5 6 7 FRICTION RATIO F.R % FIGURE 10. RELATIONSHIP BETWEEN FRICTION RATIO AND SOIL DESCRIPTION 40 30 0. o w u. o 20 0 :::E 10 Further information should be obtained to increase confidence in the chart or modify it if nece~ sary. The primary purpose described herein of measuring the friction ratio, and thus providing a description of the material type, is so that settlement calcula tions may be modified accordingly. However, there is a further important advantage in having a reliable material description. It is that a preliminary assessment of the time settlement charac teristics of the subsoil becomes possible . There is as yet no formal way in which this can be done but it is reasonable to hope that a relationship may be established between friction ratio and Cv , the coef ficient of consolidation via the plasticity index . It should be stressed that it is not the intention that sounding should replace all other testing but solely that the maximum benefit shou~d .be derived from soundings carried out so that reasonable assess ments of potential subsoil behav~our can be made at an early stage of investigation. CONCLUSIONS 1. Deep sounding is an extremely effective method for the estimation of the settlement of embank ments on alluvial saturated deposits. 2. The development of the friction jacket has con siderably increased the usefulness of the method because the resulting definition of the material type enables : a) the calculated settlement to be modified; and ·b) an assessment of the time-settlement characteris tics to be made. 3 . Measuring systems currently in operation are probably unable to provide the accuracy necessary for friction ratio measurements in very 101" densi ty subsoil. 4. An improved strain gauge system has been devised which is cheap, sufficiently accurate, convenient to use and has the very useful facility of auto matically recording the results . The equipment has been used successfully in the field . 5. Using the new equipment, friction ratios have been determined and compared with different soil parameters, these being the I , the % < 20 ~ and the % < No.200 B.S. sieve. Fo~ the sites examined a simple relationship exists between Ip and % < 20 ~ which enable the Begemann chart to be trans formed into a line on a friction ratio against Ip plot . Close agreement exists between that l~ne and the results of the field tests. 6. It is therefore concluded that a revised chart _ Figure 10 .. may be used to obtain descriptions of very low density or soft soil from the fric tion ratios, provided that the measuring system is adequate. 7. In order to increase confidence in the use of sounding friction ratios for determining material type, more sounding and associated . sampling should be carried out and the results correlated . ACKNOWLEDGEMENTS The a~thor would like to thank Mr. F. Reid and Mr. J. Vorster who designed and manufactured the im proved load measuring device and who patiently made the subsequent modifications to produce the final ver sion of it. Thanks are also due to the Director of Roads of the Natal Provincial Administration, to the Secretary of Transport and their Materials Engineers who made the sites and the field eqUipment available for much of the testing. The paper is published with the permission of the Director of the National Institute for Road Research. 175 REFERENCES BACHELlER, M. and PAREZ, L.(1965). Contribution to the study of soil compressibility by means of a cone pene trometer. Proc. 6th Int. Conf. Soil Mech. Fdn.Engng., Montreal, Vol. 2, pp. 3-7 . BEGEMANN, H.K.S. (1953) Improved method of determin ing resistance to adhesion by sounding through a loose sleeve placed behind the cone. Proc. 3rd. Int. Conf. Soil Mech. Fdn. Engng., Switzerland, Vol.l, pp. 213-217. BEGEMANN, H.K.S.(1965) The friction jacket cone as an aid in determining the soil profile. Proc.6th. Int. Conf. Soil Mech. Fdn. Engng., Montreal,Vol.l,pp.17-20. BEGE~~, H.K.S.(1969) The Dutch static penetration test with the adhesion jacket cone. Laboratorium voor Grondmechanica, Delft, Vol.12,No.4, Chapters 1 and lL DE BEER, E.E. and MARTENS, A.(1957) Method of compu tation of an upper limit for the .influence of the heterogeneity of sand layers on the settlement of bridges. Proc. 4th. Int. Conf. Soil Mech. Fdn. Engng., London, Vol.l, pp. 275-282. DE RUITER, J . (197l) Electric penetrometer for site Investigation. J. ASCE:Soil Mech. Fdn . Div., Vol.97, No . 5M2, pp. 457-472. GIELLY, J., LAREAL, P. and SANGLERAT, G.(1970) Correl ations between in-situ penetrometer tests and the compressibility characteristics of soils. Conf. on i n-situ investigations in soils and rocks. London, pp. 167-172. HEIJNEN, W.J.(1973) The Dutch cone test. Study of the shape of the electrical cone. Proc. 8th. Int. Conf. Soil Mech. Fdn. Engng., Moscow,Vol . l.l, pp. 131-134 . JONES , G.A. (1974) Methods of estimation of settlement of fills over alluvial deposits from the results of f i eld tests . NIRR unpublished report RS/6/74, Pretoria, CSIR. JONES, G.A.; LEVOY, D.F. and MCQUEEN, A.L.(1975) Em bankments on soft alluvium - settlement and stability study in Durban . Paper submitted to 6th Reg.Conf. for Africa Soil Mech. Fdn. Engng.,Durban. JOUSTRA, I.A.K. and FUGRO, N.V.(1973) New develop ments of the Dutch cone penetration test. Proc.8th. Int. Conf. Soil Mech. Fdn. Engng., Moscow, Vol. 1.1, pp . 199-201. KANTEY,B.A.(1951) Significant developments in subsur face explorations for piled foundations. Trans.S.Afr. Inst. Civ.Engrs. Vol.l, No.6, pp.159-l85. KERISEL, J.(1968) Mecanique des Sols. Paris,Dunod, p. 191. LAMBE, T.W.(1973) The 13th Rankine Lecture:Predic tions in Soil Engineering.Geotechnique,Vol.23,No.2, pp . 151-201. MEIGH,A.C. and CORBETT,B .O.(1970) A comparison of in situ measurements in a soft clay with laboratory tests, and the settlement of oil tanks.Conf.on in-si tu investigations in soils and rocks,London,pp.173~79 MEYERHOF,G.G.(1965) Shallow Foundations. J.ASCE:Soil Mech. Fdn.Div., Vol.9l, 5M2, pp. 21-31. SANGLERAT, ~(1972) The penetrometer and soil explora tion. Amsterdam, Elsevier. SCHMERTMANN,J.H.(1970) Static cone to compute static settlement over sand. J.ASCE.Soil Mech.Fdn.Div.,Vol. 96, pp. 1011-1043. WEBB, D.L.(1974) Penetration testing in ~ica. Proc.European Symp.on Penetration Testing, Stockholm. designed bearing capacity of the material exceed 4 000 kPa. It would seem that the full potential strength of the material was not used, although in some cases I realise the size of the shafts were determined for practical reasons of installation. It would be interesting to know from the authors whether smaller sections to carry the same load at the same depth would have been used had the rock " drilling equipment now available been on the market at the time of construction of the four case histor ies mentioned in the paper. On Page 163 the authors state, "High grade concrete stressed to appropriate working stresses can be safely used. In general therefore less concrete is used than in piled foundations". This would not appear to be a material factor in the economics of deep founding techniques si~ce the cost of concrete probably represents less thani10% of the total cost of installation, and with special care there is no reason why the concrete in high capacity piles should not be fully stressed as well. CONTRIBUTIONS TO THE PAPER DEEP SOUNDING - ITS VALUE AS A GENERAL INVESTIGATION TECHNIQUE WITH PARTICULAR REFERENCE TO FRICTION RATIOS AND THEIR ACCURATE DETERMINATION BY G.A. JONES. PROF V.F.B. DE MELLO There is much to be discussed in this very important session of great practical implication particularly in the light of our experiences of the last thirty years during which the principal cities of Brasil have grown at incomparable rates (such as that of Sao Paulo with an increase of popUlation from about 2 to about 8 million). This has i nvolved an unpre cedented rate of construction of highrise buildings because of disproportionately slow expansion of the area served by public utilities. However, because of lack of time I shall restrict myself to three principal items . The first concerns the paper by G.A. Jones on the use of the deep sounding static cone penetrometer together with the local friction sleeve as a preli minary all-purpose tool for subsoil investigation. When Begemann presented his suggestion of the friction ratio as "an aid in determining the soil profile" (1965), which could better have been emphasized to be a complement to visual-tactile classification of spoon-sample exploratory borings, I took the liberty to submit a discussion (6th ISSMFE , Montreal 1965, Vol. III p.294) decrying the introduction of mechani stic practices that would wipe out the painstaking gains of the fundamental principle of Soil Mechanics of requiring first the determin"ation of the nature (classfication) of the soil type" by direct sampling, and not by indirect inferences. M~ Jones begins by pitting undisturbed sampling of soft alluvial deposits against the proposed deep sounding procedure. But the latter is an Index Parameter procedure, that should be compared with the alternate index parameter procedure of the exploratory boring: the latter assumes that by visual-tactile classification one obtains principally the determinations of parameters connected with strength as the satisfactorily dominant conditioner 105 for problems both of compressibility and of stabi lity. Moreover, does one forego the need to deter mine the water level in borings? Indeed the lure of mechanistic proposals is extreme ly seductive. And doubtless there will be many a case of pragmatic success. Meanwhile, doubtless the presumed rationalizations employing the index para meters extracted from the exploratory borings do need considerable revision, to redeem them from an accumulation of criticisms on poor predictions. But, can one forego a fundamental principle of • directly qualifying the materials, without runnLng serious risks, not merely of practical failures (not yet statistically established because of the limited number of applications of the n~w method) but worse, of undermining the very roots of the engineering science? The author is much commended for the improvements introduced and the carefully collected supporting evidence. I should beg leave to request, however, that much greater emphasis be attached to the re strictive hypothese; for instance, on the one hand regarding shear strength estimations, the assumption of fully saturated type of G 2 0 soil (most partially desiccated and preconsolidated alluvia even when cyclically submerged would not satisfy the condition for a small pressure bulb under the cone), and, on the other hand regarding settlement computations, the presumed pseudo-elastic instaneous compression condition of the Buisman-de Beer- Schmertmann compu tation of sandy soils. The introduction of a new Index Parameter, such as the Modified I may improve statistical correlation coefficients; Rut, with reference to fundamental parameters and the theory of soil mechanics, is there any justification (or can and should one be sought ) for the presumed trend, or is it a case of statistics at random? It must be recalled that for routine exploration, the cone penetrometer suffers from the serious drawback of being in essence too sensitive a test, subject to highly localized extreme values. If as much effort of development and correlation were expended on the samoling exoloratory boring (such as, for instance, using a static penetration effort and a Swedish foil long-sampling idea), would not the returns to soil engineering both conceptually and pragmatically, have a probability of being much greater? The second point on which I beg leave to comment is the fact that foundations and foundation design seem to be discussed at this Conference without sufficient indication of the subsoil profile (in a soil mecha nics context) or of the magnitudes and distribution of column loadings and I find myself quite at a loss in the attempt to assess comparatively your practices with ours. If I may say so, you seem to be map conscious and geology conscious; in our experience, geology is merely the compulsory context within which to begin engineering investigation, but does not lead to any indices quantifiable to the degree required of engineering decisions, especially in urban foundation engineering concerned with restric ted areas and depths and finer property differentia tions. For instance, in paging through the Proceed ings Volume I could single out two places (pages 194, 244) in which subsoil profiles are complete for foundation engineering, with classification of soil types and consistencies, and three other places (p.84, 257, 264) in which there are subsoil profiles, but without quantification of denseness or consis tency. I \ \ since we have developed all of our routine first approximation foundation design decisions on the basis of exploratory boring with SPT indices, the boring profiles shown on pages 194 to 197 would appear to indicate conditions for economic shallow footings foundations for buildings of about 2S to 30 storeys; but one must be careful to check against possible gross differences in SPT values that can be produced by unstandardized factors of the test, leading one completely astray in comparisons (which is best avoided by comparing with ~tatic cone pene trometer values or by comparing SP! values in a standard material such as clayey fill compacted to similar specifications). I should therefore begin by requesting that a typical exploratory boring profile of Durban be published in the discussions, alongside referenced information on static cone penetrations and/or fundamental data on plate load testing, and alongside a typical plan of column loads of a highrise building of specified number of floors. , . For your reference I may summarise the following first order approximations widely used in Brasil. Reinforced concrete building loads correspond to about 1,2 tons per sq~are meter of area in plan, per floor (thus 1,2 n tim would be the equivalent average pressure on a hypothetical raft). Rafts have never proved necessary or economincal; if the allowable bearing pressure on pad footings is higher than about 1,8 n and therefore overall pad areas add up to less than 6Si. of the plan area, pads are more economical than piles of the order of 10 m; the corr.petitive limit of percentage area occupied by shallow pads continues below 1007. up to the longest piles used (about 2S to 30 m). For somewhat precon solidated silty clays and clayey sands with 3 < SPT < 25, plate load tests have suggested allow,ble bearing pressures equivalanet to ( JSPT - I )kg / cm- as being satisfactorily conservative, dispensing with computa tions of settlements and differential settlements on routine range of variation of column loads. Pier foundations have accepted nominal base pressures (assuming zero friction) of the order of 2 to 3 times the above shallow pad indications. For driven displacement piles, the pile length necessary to permit the design load is estimated on the basis of SPT values along the boring profile, with one value per meter of depth; if the layers are considered capable of concomitant contribution to friction and point, the leng~h is such that E SPT = compressive stress in kg/cm on the nominal concrete section (for instance, a 30 x 30 cm pile for 40 tons would penetrate to about ESPT = 45); if there is a signi ficant distinction between soft upper layers and embedment into a dense substratum of point resistance, the substratum will be penetrated to where ESPT point = one-half the compresive stress. Such secret unwritten rules are denied as soon as they are passed along, but have served for most preliminary designs; and since in Brasil most often construction follows rapidly upon preliminary design they may be claimed to have been proven. I submit them merely so that they may be compared, challenged, and put to shame. However, in candid contrast may I ques~ion whether the assumptions of c and 0 paraQleters for "upper strata" and "Cretaceous material" in the pa~er by Everett and McMillan are not merely cloaked w~th an appearance of acceptable theorization· the crucial problem of pile or pier foundation ' design is "how were these parameters established"? 106 Finally, the third point concerns the all-important "execution effects" both for lateral friction and for stress-strain behaviour of the concreted base. In particular, the effects of bentonite (and pre sumed bentonite cakes) on skin friction and on base compressions have drawn much attention and testing in the past few years (cf. for instance some papers at the European SMFE conference, Madrid 1973). The papers by Everett and McMillan, and by Wates and Knight, tackle a problem of the greatest interest and concern. In Brasil, and particularly in the kilometers of deep slurry walls of the Sao Paulo and Rio subways we have had considerable success in general; but one MuS t carefully guard agians t the new "philosop her's stone" complex. Each case must be examined separately; bentonite is not a cure-all, and there are many cases where it is unnecessary or may even be damaging. Of special note to this Conference is the admonition that in soils above the water table a "dry" perforation technique that preserves the benefits of capillary tension may be very much better, since despite the best of bentonite slurries the contact with free water frequently causes cata strophic damage to the soil. DR A. P. TYRRELL In the concluding paragraph of the section in Mr Jones' paper entitled 'Interpretation of friction ratios' it is stated that "it is reasonable to hope that a relationship may be established between friction ratio and ....... the coefficient of con- solidation ....... ". I wish to challenge this statement for it is totally ~-reasonable to have such a hope. ·c In general terms, the deep sounding technique is a useful tool, but it should be realised that it has severe limitations. For alluvial ·clays, for in stance, it provides only a very superficial means of identification. There is no substitute for visual examination of these clays which have been found to exhibit a natural fabric (e.g. fine sand and silt layers and partings, rootlets, etc.,) capable of dominating mass performance in the field. In order to be able to make realistic engineering predictions, therefore, it is essential to identify this soil fabric and then to appreciate its influence on mass behaviour (Tyrrell 1969). Descriptions such as those shown in Figure lOot the paper viz., 'clay', silty clay' can be entirely misleading with regard to field rates of consolidation (Rowe, 1968) and highlight the limitation of the deep sounding technique. REFERENCES: ROWE, .' P. W. 1968. The influence of geological features of clay deposits on the design and per formance of sand drains. Proc. of the Institution of Civil Engineers, Supplementary Volume, Paper 7058S. TYRRELL, A.P. 1969. Consolidation properties of composite Soil deposits. Ph.D. Thesis, University of Manchester. DR B.C. VAN WYK The first Dutch cone apparatus had a conical point which was attached to the internal pushing rods. The main objection to this point was the sharp drop in point resistance observed when the soil broke . in behind the cone. The operators had to be attent4ve or they could miss the maximum point resistance reading. Subsequently the jacket cone penetrometer was introduced and this gave a more continuous resistance and the maximum value was easier to determine. Some engineers w~re reluctant to use the jacket cone as it apparently gave higher point resistances. The new friction sleeve cones are again geometrically different from the previous cones. I would like to describe the apparatus tes ted by the Delft Soil Mechanics Laboratory (Heijne.n 1973). It was found that this apparatus gives the best agreement with the jacket cone. It had an 28 rom narrowed part, 200 mm long directly behind the conical point with a 15 000 sq mm friction sleeve at a distance of 300 rom behind the ·cone point. The author also mentioned thntithe settlements predicted with the compressiod modulus calculated from the point resistance of the Dutch cone was found to be on the high side. Quite a number of factors contribute to this discrepency and I would like to comment on one aspect. Tests carried out under laboratory conditions in large containers on clean homogeneous sands by Kerisel (1964) gave a particular penetration diagram. Over the first 15 diameters there is a sharp in crease of penetration resistance with depth which is attributed to the development of the complete deep foundation failure pattern, which, incidently, has not been recorded by any investigator so far. After this the penetration diagram bends down and becomes pratically constant with depth at about 30 diameters. Wi th the overburd'en pressure increasing with depth and the point resistance constant, this implies that the material is becoming more compressible with depth, and that under an increasing normal stress. The picture of a dense layer under the overburden of a soft layer follows from the work of Thomas (1968). He carried out penetration tests in a rather small container with the effect of a surcharge simulated by a water pressure on a latex membrane and a steel plate. For a constant surcharge pressure the re sulting penetration resistance was constant with depth from the surface down. The penetration resis tance increased with increasing surcharge pressure but tended to reach a maximum at about 550 kPa surcharge or the equivalent of 4 m of overburden. This ties in with the work of Kerisel. It also implies an increase of compressibility with depth after about 4 m. References Heijnen, W.J. (1973). The Dutch Cone Test. Study of the shape of the electrical cone. 8th Int. Conf. Soil Mech. and Found. Eng. Moscow. Vol. 1.1 p 181. Kerisel, J. (1964). Deep foundations basic experi mental facts. Deep foundations conference. Mexico. Thomas, D. (1968). Deep sounding test results and settlement of spread footings on normally consoli dated sands. Geotechnique 18; p. 479. K. SCHWARTZ In reading this paper, coupled with a paper by the Author and others reported in Session 5 of the Conference, it appears as if some success has been achieved in predicting settlements for road embank ments using deep sounding test results with the modi fied Buisman-de Beer method of settlement calculation~ This method, proposed by de beer and Marstens (1957), is based on the standard Terzaghi settlement equation with the relevant relationships being given in equations 3 and 5 of the paper. These equations have in general been considered valid for saturated loose sands in which the stress incremF.nt due to the applied load is small compared with the overburden pressure, and in which Boussinesq's theory of stress distribu tion is valid. In addition, in assessing settlements it is necessary to distinguish between normally consolidated and over consolidated deposits. During the field work stage of recent projects it has been necessary to carry out preliminary settlement predictions using deep soundings in saturated cohesive clayey silt soils. In these calculations the Buisman de Beer relationship for the calculation of the compressibility modulus in saturated loose soils has been modified as follows, in an attempt to take the soil type into consideration. 107 ----------- Equation 1 is modified to C ----------- · Equation 2 where a ----------- Equation 3 The methods proposed by Gielly et al (1970) have been used to estimate values for ao ' The values of a o vary considerably ' with the soil type and have a direct influence on settlement calculations. No field corre lations are available at present to check on the validity of the above assumptions. The Author has, however, carried out similar investigations in cohe sive soils on the Sea Cow Lake site. In the settle ment calculations reported by the Author and others in a paper presentc~ in Session 5, correlations have been made with field measurements. The predictions using deep sounding and the field measurements correlate reasonably well, even although Equation 1 above has not been modified to take into account the fact that the deep soundings have been carried out in sandy or silty clays. The validity of using an ao factor as given in Equation 3 above is questioned, and further comments by the Author would be appreciated. The Author's work on correlations between friction ratios and material type is most interesting and could become a useful tool in site investigation if the reliability of the method could be proved. If one considers the friction ratios obtained from deep soundings on the Mtwalumi site as given in Figure 3 of the paper, in association with Figure 10, one obtains variations in soil type from a clean sand to a clay over very small vertical distances. One queries the reliability of the method when the Author states that boreholes drilled on the site gave a soil profile of 25 m of silty sand overlying Dwyke Tillite Bedrock. References i. DE BEER, E.E. and MARSTENS, A. (1957). Method of computation of an upper limit for the influence of the heterogenity of sand layers in the settlement of bridges. Proc. 4th Int. ConL Soil Mech. Edn. Eng. London. ii. GIELLY, J., LAREAL, P. and SONGLERAT, G. Correlations between in-situ penetrometer tests and the compressibility characteristics of soils. Conf. on in-situ inv~stigation in solids and rocks. London. AUTHOR'S REPLY: The paper on Deep Sounding may be divided into three sections: I. Prediction of settlement of embankments from sounding results. 2. A description of some modi~ications to the conventional load measuring devices. 3. The use of friction ratio measurements obtained from the mechanical friction sleeve cone. The first of these, embankment settlement, will be discussed in greater detail in the session on em bankments; however, I would like to mention that this paper gives genuine predictions for settle ments of embankments before they were constructed, a practice which I am happy to see Dr Burland advocates. The second section deals with a fairly simple modi fication which we, at the National Institute for Road Research, have made to the usual load sensing device for probing. -This has advantages over the usual hyrau1ic load cell and gauges in that i: is more accurate and that a permanent record is made of the output on a chart recorder. It is appreciated that very much more sophisticated devices are avail able but in general they are expensive and relative ly complicated. The third section has the most relevance to piling and that concerns the use of friction ratios mea sured by the mechanical friction sleeve in the construction of a soil profile. What we have done certainly makes no claim to be original - we have simply confirmed that for some recent alluvial deposits in this country the conventional friction ratio interpretation is applicable. We have found it convenient to express the results in a somewhat different manner and give a simplified chart of friction ratio against soil type. The relevance of this to piling is of course primarily for friction piles where a knowledge of soil type is essential for meaningful calculations to be made. The book by Sanglerat is vital reading on this subject. Generally the main point being made is that sounding is a very convenient tool, backed by years of expe rience elsewhere, and we should be making more use of it. On the other hand, the last thing I want to do is to suggest that sounding is the only, or most important, technique which should be used, or that som~ of the semi empirical factors associated with interpretation should be viewed with mystical sanctity. 108 CONTRIBUTIONS TO THE PAPER TIME EFFECTS ON THE LOAD CARRYING CAPACITY OF DRIVEN PILES IN THE DURBAN AREA BY S.B. SHARRATT DR W. NEELY The Author has presented some interesting data on the increase in pile bearing capacity as a result of redriving at various intervals of time after initial driving. The prediction of pile capacity was based on the Hiley pile driving formula. This procedure or indeed any method of load prediction based on dynamics will only give the pile capacity immediately after driving. In normally consolidated clays the pile capacity usually increases after driving giving rise to the commonly observed phenomenon of 'set up'. In overconsolidated clays and sands the pile capacity may decrease with time. Any appreciation of the factors which are responsible for increases in pile capacity with time may have an important commercial value in addition to providing a more meaningful guide to the time lapse between driving and testing in order to take advantage of the 'set up' behaviour. In Norway and Sweden, for example, it is established practice to test piles no sooner than a month after driving, Bjerrum, Hansen and Sevaldson (1958) have found that this is suffi cient time for a pile to reach maximum capacity. It has been well known for many years that pile driving results in an increase in porewater pressures in the surrounding soil which are then dissipated after driving by horizontal drainage. The rate of drainage is clearly controlled by the permeability of the soil although the pile properties may also have a significant influence. The porewater pressures created by pile driving can be estimated by considering the soil displacement around the pile as illustrated in Fig. I. Let the initial in-situ vertical and horizontal effective stresses be J-vi and KoJ-vi respectively and the porewater pressure prior to driving be ui. During driving the direction of maximum displacement is radially outwards and therefore within the zone of radius R\ the radial stress becomes the major princi pal stress. According to La and Stermac (1965) the maximum porewater pressure, 6 U m , is comprised of two components as follows: 6- U a (I-K o ) rvi (due to change in 6J3) 6 Us ( 6 U/P)m J-viCdue to shearing) The maximum porewater pressure produced during driving can then be found provided the coefficient of earth pressure at rest, K , and the maximum pore pressure ration l6 U/P) canobe measured. (6 U/P) can be determined from ~onso1idated undrained tria~ial tests with porewater pressure measurements while K can be estimated for normally consolidated clays a from the work of Kenney (1959). The distribution of porewater pressures around a pile is usually assumed to vary according to the relationship shown in Fig. 2. UR R UM ~ This distribution has been substantiated by field measurements of porewater pressures around driven piles, e.g. Hanna (1967). The rate of dissipation (ii) PREDICTION OF TIME FOR CONSOLIDATION FROM SOUNDING 1/28 Prediction of Time for Consolidation from Sounding Prediction de la Duree de Consolidation des Materiaux Provenant de Sondages G.A.JONES Soil Engineering, National Institute for Transport and Road Research, South Africa SYNOPSIS For road embankments over alluvial deposits the time taken for settlement is often of greater signi- ficance than the magnitude of settlement. A method of estimating the time, based on deep sounding, is proposed in which the test is carried out at a constant stress . In order to demonstrate the feasibility of the proposal, a series of laboratory tests was carried out comparing consolidation characteristics measured by constant stress penetration , with those measured by conventional consolidometer tests. INTRODUCTION The planning of routes for highways requires prelimi nary geotechnical investigations. Estimates of the amount of settlement of embankments and, equally im portant, the period of settlement are often required. For such investigations an indication of the order o f magnitude i.e. 0,1; 1 or 10 years may well be suffi cient. I n South Africa, quasi static penetrometer testing (deep sounding) is used almost on a routi ne basi s for estimating the settlement of embankments on alluvial deposits (Jones 1975, webb 1974). Although di scus sions about the theoretical interpretation of penetra tion testing data are continuing there is little doubt that the use of the technique on a semi empirica l basis is justified. At present the drawback of sound ing is that it gives no indication of the duration of set-::.lement. It is here reported that estimates of the ~equired consolidation characteristics were made from penetra tion tests carried out at a constant stress , instead of at the more usual constant rate of penetration, on a series of laboratory samples on which consolidometer tests were also carried out. TEST RESULTS A standard mechanical friction cone penetrometer was mounted vertically in a frame as shown in Fig. 1. A loading platform was attached to an extended inner rod and an LVDT connected to a chart recorder was arranged to measure the movement of the platform. Soil samples were prepared in an inner bucket surrounded by a water jacket. The inner bucket had perforated sides and was lined with a filter fabric. The- frame was designed so that the penetrometer could be moved to different d~pths in the bucket. In this way a number of tests could be carried out on the same sample. A piezometer was fitted into the head of the cone for some of the tests. It consi"sted of a pressure transducer located inside a cone with porous stone windows in the face. The procedure was analogous to consolidometer testing in some respects. Loads were added to the platform and a series of graphs of deformation as a funct i on of time was obtained. On completion of a test the cone was lowered to a new position and a further test car ried out. Three soil types were used, silty sand, clayey sand and a silty clay. The first two samples were prepared in the bucket whereas the clay was an undi sturbed sample extracted by an excavator from a site currently being utilized for embankment studies (Jones et al 1975). Fig. 1 Constant stress penetrometer The claSSification test results on the three soils are " given below: Sample i Particle Size Distri- Atterbcrg . Description bution 'I Limits > 60u <60u >2u < 2u WL Wp Silty clay 13 35 52 56 35 Clayey sand 55 22 23 33 20 Silty sand I 93 7 - N.P The results of standard colsolidation tests on the three samples are shown on Fig. 2. The results of the deformation-versus-time penetro meter tests are given in Fig. 3. The tests on the clay showed evidence of what may be regarded as secon dary consolidation. The end of primary deformation 135 1/28 was graphically determined for each material and all deformations were expressed as a percentage of the se and plotted against time on a logarithmic scale. Pore pressures were also recorded during some of the tests on the clay samples. .. z II ... .. :> '" ~ o Fig. 2 TIME MINUTE S Consolidometer deformation vs. time 100 o~------....L.-"":::"'------',o-----~---'o()o TIME MINUTES Fig. 3 Penetrometer deformation vs . time DISCUSSION The consolidation tests results for each material t ype see Fig.2, show fairly narrow bands whereas the pene trometer results in Fig. 3 show much wider spreads . The pore pressures rose immediately on application of a load after which there was a gradual decay apparent ly consistent with the decreasing rate of penetration. However, as is well known, such pore pressure measure ments can give rise to problems of interpretation (Holden,1974, Scnmertmann, 1974) although many have advocated their use (Hansbo,1974, Janbu and Senneset, 1974, Ladanyi,1976). A comparison of Figs. 2 and 3 shows that consolidometer and penetrometer test times tend to agree although there are clearly anomalies which require explanation. In order to use the pro posed penetrometer test with any confidence it will be necessary to resolve these anomalies and to collect considerably more data . It was observed that in all the materials the shape of the deformation time plot could be significantly altered by changing the load. This was presumably because the materials failed at higher stresses with the result that consolidation effects were masked. In this preliminary series of tests the maximum deformation was not controlled by limiting the loads and it was arbitrarily decided to discard results where the deformation exceeded 10 per cent of the cone diameter. It was accepted that the laboratory tests indicated sufficient correlation between the consolidometer and penetrometer results 136 to justify field testing. This was found to be straightforward if the uppermost inner sounding rod was fitted a load platform . A standard 10-ton Goudsche Machinefabriek p~obe was used which had been fitted with an electrical strain gauge load measuring system with a chart recording device (Jones,1975) . Deformation measurements were taken with an LVDT con nected to this system. CONCLUSIONS Constant stress penetrometer tests are considered to~ feasible for the preliminary field estimation of the time-settlement characteristics of alluvial deposits . Just as conventional constant rate of penetration testing has required a great deal of field correlation so will constant stress tests require similar correl ations with other time-dependent tests and field performance to prove their validity. This paper is published by permission of the Director of the National Institute for Transport and Road Research. The author also wishes to thank Messrs van Loggerenberg and Vorster of the Institute for their assistance with the design and construction of the equipment and carrying out the tests. REFERENCES HANSBO, S. (1974), "Pore pressure sounding apparatus," Proc. European Symp. on Penetration Testing (ESOPT) , Vol.2:1, pp. 109-110 . HOLDEN, J.C. (197 4), Gene ral discussion, Proc . ESOPT , Vol. 2 : 1, pp. 100-107. JANBU, N. and SENNESET, K . (1974), "Effective stress interpretation of in-situ static penetration tests". Proc. ESOPT, Vol. 2:2, pp . 181-193. JONES, G.A. (1 975), "Deep Sounding - its value as a general investigation t echnique." Proc. 6th Reg. Conf . for Africa S.M.F.E., Vol.1, pp. 167-175. JONES, G.A., LEVOY, D. F. and McQUEEN, A.L. (197 5 ), "Embankments on soft alluviam - settlement and sta bility study at Durban," Proc. 6th Reg. Conf. fo r Africa S . M.F.E., Vol. 1, pp. 243-250 and Vol . 2 , pp . 134-137. LADANYI, s, (1976), "Use of the static penetratiop. test in frozen soils," Can. Geotech. J., 13 pp. 95- 110. SCHMERTMANN, J . H. (1974), "Penetration pore pressure effects on quasi-static cone bearing," Proc. ESOPT, Vol . 2:2, pp. 345-351. SCHMERTMANN, J.H. (1974), General Discussion. Proc . ESOPT, Vol. 2:1, pp . 146-150. WEBB, D.L. (1974), "Penetration testing in South Africa," Proc. ESOPT, Vol. 1, pp. 201-215. (iii) EMBANKMENTS ON SOFT ALLUVIUM SETTLEMENT AND STABILITY STUDY IN DURBAN · Sixrh R~gional Conference for Africa on SOIL MECHANICS & FOUNDA TlON ENGINEERING Durban, Sourh Afric~, Seprember 1975 Embankments on . soft alluvium -settlement and·' stability study at Durban G.,;\,. JONES Chie{ Research O{ficer. National Institute {or Road Research. CSIR. Pretoria D ~ F.: LE YOY Formerly partner o{ Saunderson. Le Voy and Partners. Durban A.L. MCQUEEN Partner o{ A.A. Loudon and Partners. Durban SYNOPSIS A series of road embankments is to be constructed over an area of soft alluvial deposits. Prelimi nary 1nvestigations cons\sting of boreholes and deep soundings have shown that settlement and stability pro blems are likely. The proposed construction programme requires reliable predictions of the time-settlement characteristics and since conventional investigation cannot given these, a trial embankment was built. This was comprehensively instrumented with settlement sensors and piezometers. The measured settlement of the 6 m high embankment was 1,3 m which agreed well with a prediction made from field and laboratory test results. The time in which this settlement occurred however was only about 500 days compared with the predicted time of 100 years. The trial embankment showed that the construction programme could be adhered to without recourse to expensive drainage measures. A contract has been awarded for earthworks only and the pavement layers will be constructed two years later. Since the stability analyses have shown that the proposed embankments will be only marginally stable, permeable blankets and toe berms will be incorporated and construction will be monitored with piezometers. RESUME On a t!tabli un proj et de construction d 'une st!rie de remblais routiers sur des gisements alluvion.- naires tendres. Une t!tude prt!alable, effectu~e ~ l'aide des trous de forage et du sondage ~ grande profondeur, a mis en ~vidence la possibilit~ des probl~mes de tassement et de stabilit~. Le programme de construction exige des predictions stires pour des caracteristiques temps-tassement. Puisque les etudes classiques ne peuvent pas les fournir, on a construit un remblais a titre d'essai, muni de l'appareillage complet comprenant les senseurs de tassement et les piezometres. La mesure d'un remblais de 6 m d'hauteur a montre un tassement de 1,3 m, qui s'accordait bien avec la prediction basee sur les resultats des experiences de laboratoire et de chantier. Le tassement n'a pourtant dure que 500 jours environ, tandis que la prediction estimait une centaine d'annees. Le comportement du remblais d'essai a demontre que Ie programme de construction peut !tre poursuivi sans que des mesures de drainage couteuses soient necessaires. On a conclu un contrat uniquement pour les tra vaux de terre, tandis que les couches du corps de chaussee ne seront posees qU'apres deux ans. Puisque l'analyse n'a demontre qu'une stabilite marginale des remblais projetes, on va incorporer des tapis permeables et des pieds de banquette, et de plus controler les travaux de construction a l'aide de piezometres. INTRODUCTION This paper describes an investigation, with stability and settlement analyses, which was carried out for a series of embankments to be constructed over compressible alluvium. The embankments carry part of the Durban Outer Ring Road which is shown in Figure 1. This section of the road under considera tion runs from the umgeni River northwards for about 12 km; the embankment problems occur in the southern 6 km, (i.e. chg 6S0-chg 830 in 100 ft chains). The route was located to minimize the expropri ation of domestic property, with the result that some considerable geotechnical problems have arisen through following a winding valley in a faulted area. The valley is low-lying and subject to frequent flooding. The meandering of the river, the Umhlan gane, has resulted in four separate crossings within a distance of about 4 km. 243 Even along the parts of the proposed road which are not problematic because ·of swamps or faulting, other geotechnical difficulties occur. Although not the subject of this paper, it is interesting to draw attention to one of these parts as a good example of planning coordinated with site investigation in its various phases. In one place the preliminary road line passed along the east side of a hill in a small side cut and by so doing avoided a swamp. However, the engineering geologist responsible for the soil engineering map had noted, from interpretation of the aerial photographs, that the locally well-known pro blem of east dipping Ecca shale was likely to be present. A simple site inspection of the small cut tings for domestic driveways confirmed a very marked dip practically at right angles to the road line. An assessment of the situation showed that it was preferable to realign the road through the swamp area rather than incur problems of stabilising even quite small cuttings. FIGURE 1 SITE PLAN PRELIMINARY INVESTIGATION A considerable amount of preliminary site inves tigation was carried out along the route. Apart from the soil engineering mapping, this consisted of sink ing a number of boreholes and carrying out standard penetration tests (SPT), and later a series of deep soundings. Table 1 gives an indication of the poor subsoil conditions by listing the SPT values against depth for typical boreholes, which are numbered by their chainages along the centre line. As a follow-up to the boreholes with SPT's. 35 soundings were carried out in the swamp areas at in tervals along the centre line. During the course of this work, in-situ vane shear tests were performed so that a correlation, for the site, of the undrained vane shear strength, cu, with the deep sounding cone pressure, qc' could be made. This is discussed elsewhere Jones (1975). Figure 2 shows the resultant relationship, together with a typical borehole log. Factors of safety for circular arc failures could then be evaluated using assumed fill parameters and subsoil strengths, estimated from deep soundings . The Krugmann and Krizek (1973) charts and Pi1o~ and Moreau (1973) charts were used. Since the shear strength of the fill material was assumed, the esti mates could only be considered as a useful prelimina ry assessment. The settlement of the proposed embankments was estimated from the sounding results by using the Buisman-de Beer relationship (de Beer and Martens .(1957) in conjunction with the Terzaghi consolidation 244 equation: 6H where ISH H Po Oz C and '/c where qc Depth m 1,5 3,0 4,5 6,0 7,5 9,0 10 , 5 12,0 13 , 5 15,0 o 5 E x .... "- '" 0 10 ~':,( ~~ 15 2,3 x H C settlement of layer thickness of layer (1) overburden pressure at mid depth of layer increase in stress at mid depth of layer due to load compression modulus Po 1,5 qc (2) = deep sounding cone pressure. I GWL .3?- TABLE 1 SPT AGAINST DEPTH Borehole No. Chainage (100 feet) 666 682 1 0 7 0 0 7 5 47 59 65 CARK GREY BROWN SOFT CLAvEy TOPSOIL WITH ROCkS 'Ieqy OARk GAEY IIEAy SOFT $l.IGHfl,..'I' S~Oy SllT"'r CLAY DARI( GREy SOFT SILTy CLAY BLACK VEAY SOF'T SANDY SILTY ClAY DARK GRE Y MOWN SOFT vERY SANOY SILTY CLA" a.llhC GA(v eROWN lOOS( CLAVEY SILTy SAHD BLACK W(ATMEIiIED SOFT 688 736 0 5 8 7 0 1 4 1 3 36 t I 0 0 0 0 8 28 _t··L~~ · --- t · ~- ' . .. t -·~ t t ~ . - - j ~ ::T ·~.-·~= I· ·t ·· · · ··. ·· • ... - - ;l ~- :-·t - t- .-:: -. J...~. := - r-~ - -- t·· T- +---..., '/1 .' ~ fl . . . .t..~ I i -!. t-.. . -I- - . ...L . . • _-1.--::J '"1iI.CluREO L4 ......... T£0 SHALE 0 N <".I CONE RESISTANCE MPo TYPICAL FIGURE 2 BOREHOLE AND SOUNDING LOGS SHOWING VANE SHEAR-CONE RESISTANCE COMPARISON • Charts were drawn up to show the predicted set tlement of an embankment for various depths of com pressible subsoil, heights of embankment and cone pressures. Figure 3 illustrates two of the chart~ . . The above calculations take no account of. the var~ab~­ lity of the subsoil but nevertheless prov~de a gu~de for estimating purposes. 1,2 SUBSOIl. DEPTH 12 m 1,0 0,8 E a 1,2 1,0 O,B E .... z . W ::E w ...J .... .... "J Vl 0,2 o CONE 2 FIl.l. HEIGHT m FIl.l. HEIGHT 8 m 8 DEPTH m FIGURE 3 12 RESISTANCE, FILL HEIGHT, SUBSOIL DEPTH AND SETTLEMENT RELATIONSHIP Various modifications to the above simple equa tion have been proposed by a number of investigators and are mentioned elsewhere (Jones 1975). The results of the preliminary investigation clearly demonstrated that stability and settlement problems were to be expected. The time-settlement characteristics could only be assessed from a descrip tion of the boreholes samples. Since these were predominantly medium to dark grey, slightly sandy, silty clay it was anticipated from local experience that the large settlements predicted would take about 2 years to reach 80% of the ultimate settlement (see Figure 4). This may have been acceptable for a continuous embankment, but since there were to be numerous bridges, problems of differential settlemen~ and possibly negative skin friction effects on the bridge piles, were important. The overall construc tion programme was therefore altered so that along this section of the route an early contract, compri sing earthworks only, was let. Less time was then available for the extended subsoil investigation than had originally been esti- mated, so that it became necessary to begin this second phase immediately. For this reason, and, also because the problem extended over a seri:s of embank ments totalling about 1,5 km in length, ~t was not possible to carry out a detailed investigation for the whole site. . It was therefore decided that a localised area should be chosen for the detailed investigation and the results extrapolated to the remainder using deep sounding values and basic subsoil parameters taken from laboratory tests of samples obtained from an auger survey. 245 ~--~r7--~-------------------- TINE - O.n5 FIGURE 4 PREDICTED AND ~IEASURED SETTLHIENTS OF TRIAL EMBANKMENT DETAILED INVESTIGATION Equipment and expertise were not available to carry out the highly sophisticated sampling and test ing procedures advocated by Rowe (1971) involving the use of large diameter thin-walled piston samplers. Of necessity, conventional U 102 sampling (undistur bed 102 mm diameter open drive tubes) formed the basis of the investigation procedure. It is now widely accepted that estimates of the time-settlement characteristics of recent alluvial deposits from conventional consolidometer testing are subject to very large errors. Closer estim~tes are possible from Rowe Cell consolidometer testing, which permits horizontal drainage of the samples (Rowe and Barden 1966), and also from the results of in-situ permeability tests. However, in view of the proposed overall construction programme, even these methods were not thought to be accurate enough for the pre diction necessary at the site. It was therefore decided that a full-scale trial embankment would be the most reliable method of obtaining the required data. Description of trial embankment site The site for the trial embankment was chosen, from the preliminary investigations by boreholes and soundings, as one of the areas with the poorest sub soil conditions. In addition it wa~ very close to a proposed large cutting which was intended to supply much of the embankment material. The position of the site is indicated in Figure 1. Although the trial embankment was to be con structed rapidly without any density control and would therefore not be included la~er in the perma nent works, it was thought worthwhile to build it on the proposed centre line. In this way one of the worst areas would have been subjected to considerable preloading. Figure 5 shows a cross-section of the 30 m by 20 m by 6 m high trial embankment. The low n:atural ground-level should be noted: it was respon sible for the frequent flooding and the consequent delay in installing the instrumentation. LEvEL GAuGES .1. PIEZOMETERS • 15m E ____ ~ ____ .::- ___ = ____ J.. ____ ~ ___ _ ..J W • > 0 w • ..J • • • • • • a • w - 5 u • • • • • • ::J a • UJ a:: -I • • • • • • • • • • • • A B C 0 E F G -15 FIGURE 5 CROSS SECTION THROUGH TRIAL E~IBANKMENT AND SUBSOIL SHOWING PIEZOMETERS AND SETTLEMENT SENSORS Boreholes, with Ul02 sampling, and deep sound ings were put down at each corner of the embankment. The boreholes were marked NE, SE, SW and NW. A typical borehole log is shown in Figure 2. Natural moisture contents, plasticity data and the results of particle size analyses are given in Table 2. TABLE 2 Depth Natural Liquid Plastic Clay Silt m m.c. % limit Limit % % <2\.1 6011 >%>211 % 2 57 56 35 52 35 4 50 66 24 45 42 6 93 70 27 67 30 8 91 63 35 60 32 10 72 69 32 45 41 12 57 60 30 49 48 14 28 30 15 29 32 Instrumentation of trial embankment As shown in Figure 5, 28 pneumatic piezometers ~nd 7 settlement indicators were installed in seven vertical profiles down to a depth of about 15 m. The piezometers were installed at vertical inte~ vals of approximately 3 m in 150 mm-diameter bore holes. Each instrument was placed in the centre of a column of sand about 2 m high and each column was separated from the sand above and below by layers of the local clay and rammed bentonite balls. Since four piezomet7rs were placed in each borehole, consi derable practlcal problems were encountered during the installation and sealing of the upper ones 246 because of interference from the read-out tubes. However all but one of the piezometers gave readings commensurate with the installation depths. Both the settlement gauges and the piezometers were commercially available ': Terra Technology Settle ment Sensor Model S-60l0C and Terra Technology Pneu matic Piezometer Model Pl020. The read-out for the instruments was obtained by measuring the air pres sure required to operate a check valve in the instru ment. In the case of the piezometer the valve balan ced the applied air pressure against a diaphragm which sensed the pore pressure through a filter element. The settlement indicator operated by balan cing the applied pneumatic pressure against a head of mercury in a closed tube leading from the indica tor to a resevoir in the gauge house. The pneumati~ pressure was applied by releasing air from a portabl~ compressed-air bottle carried in a case also contain ing the necessary pressure gauges, (control Unit Model C-6300). Connection was made to each instru ment in turn by quick-connect adaptors. All the tubes were led to a gauge house as shown in Figure 5. The gauge house level was in turn rela ted to a fixed bench mark installed on a nearby hill side. A feature of the instrumentation was the ease and rapidity with which measurements could be taken. It was found that with only a few minute's practice the reading of one instrument could be accomplished in as little as 1 minute, including the connecting up time. This point is stressed because the cost of instrumentation is sometimes erroneously taken as being simply the cost of the instruments plus that of installation. Actual costing, which should include the reading time, frequently shows that the latter is the most expensive item, particularly for a long-term project. An additional measuring system was installed at a later stage : this consisted of a large number of accurately surveyed pegs in rows parallel to the toe on three sides of the embankment. The purpose of these pegs was to measure horizontal and vertical movement at natural ground-level adjacent to the toe since at one time after the fill had been completed it was thought that the rate of settlement was in creasing . It was surmised after much puzzlement that some significant deformation may have been the cause but it subsequently transpired that, for a few of ' .the readings, the settlement of the gauge house it self had been omitted from the calculations, thus significantly altering the shape of part of the time-settlement plot! Although very large overall fill settlement was measured, no significant hori zontal movement of the pegs was detected. However, they were only monitored for a month before the tem porary omission of the gauge house settlement was rectified, so that no conclusion can be drawn. In retrospect it seems unfortunate that more use was not made of this simple measuring system throughout the whole of the trial embankment period . . A further ~oint to be borne in mind with regard to ~nstrumentatlon schemes is the necessity for maklng them vandal-proof. Despite the fact that the read-out pipes were contained in ducts leading into th7 gauge house (a l~cked shed which was guarded) thleves managed to dlg up the ducts practically at ground-level where they entered the shed, and remo ved a length of the plastic tubing from each instru ment. It was not possible to rejoin the mercury filled. leads to the settlement gauges. Fortunately, by addlng furt~er. leng~hs of tube to the piezometer l7ads and contlnulng WIth the readings, it was pos slble to 70mpare before and after readings and thus deduce WhICh tubes belonged to which piezometer_ Settlement readings were continued by placing pegs on top of the completed fill and levelling these in the conventional way with reference to the fixed bench mark. Typical readings from- these instruments are shown in Figures 4 and 6 . N ~ Z ... C> z 6 '" w a:: 50 a:: ..... 0- W ~ o N W Ii: ~ 2 w CJ) '" w a:: u z 7,5 DEPTH OF PIEZOMETER IN m SHOWN ON CURVES --- -- -- --- TIME DAYS FIGURE 6 TYPICAL PIEZO~~TER READINGS Laboratory Testing Samples were obtained from the four corner bore holes of the trial embankment and from trial pits. An auger survey was carried out along the centre line of the proposed road to obtain samples for comparative testing for the other embankment positions. Laboratory tests consisted of quick, undrained triaxial tests, drained triaxial tests, shear box tests, conventional consolidometer tests, ind i cator tests and hydrometer analyses. Tables 3, 4 and 5 summarize the results of the labo ratory testing programme. a) Quick, undrained triaxial tests. (~ 0) TABLE 3 BH No Depth Cu m kPa SW 4 21,7 SW 8 16,1 SW 12,3 7,0 NW 4 26,6 NW 8 18,2 b) Drained triaxial (T) and shear box (S) tests 247 TABLE 4 Depth Test c' V m type kPa 6 T 0 26 8,5 T 0 20 2 S 0 24 2 S 0 23,5 8,5 S 1 19,5 12 S 0 33 c) Consolidometer tests Ten conventional consolidometer tests were car ried out and the results are summarized in Table 5 . Figure 7 gives the settlement vs log-time plot ~t 50 kPa for the sample from BH. No. NE at 4,0 m. This is typical of the results, as is the void ratio vs log-pressure plot given in Figure 8. TABLE 5 BH Depth Mv Preconso- tso t90 Cv 1'l0. m (m2/kN) lidation (mins) (mins) (m2/ yr) Pressure kN/m2 ~E 2,0 0,00126 26 ISO 0,40 ~E 4,0 0 , 00041 140 9 120 1,14 ~E 4,0 0,00054 130 20 170 0,52 INE 6,0 0,00094 36 225 0,29 INE 8 , 5 0,00172 90 23 220 0,45 NE 8,5 0,00176 80 24 215 0,43 NE 11,0 0 , 00015 8 79 1,29 SE 2,0 0,00076 45 210 0,23 SE 6,0 0,00103 27 195 0,38 SE 10,0 0,00055 20 118 0,52 In the above table ~ is quoted for the 100-200 kPa pressure range and tso, t90 and Cv for the 200 kPa pressure increment. The preconsolidation pressures are estimated from the e-log p curves using Casa grande's construction. While many of the test data are still to be analysed, the indications are that the subsoil is normally consolidated at depth but that from about 4 to 5 m there is a stratum of over consolidated material. At this depth a marked in crease in shear strength is consistently shown by both the soundings and the in-situ vane tests. It may also be noteworthy that the relationship between these, which -is shown in Figure 2, indicates the largest divergence from the mean value of 18,4 in this stratum. d) Indicator tests and particle size distribution . Table 2, given earlier, summarizes typical data from the above tests for samples from Borehole No.SE . About 120 indicator tests were carried out on the samples from the auger survey and plotted on a long section so that the poorest subsoil areas could be readily detected and compared with the trial embank ment site. BH No NE DEPTH 4 PRESSURE 50 kPo : - "'!!!->-~ • 'Q .. .!: z o S o J o en z o u I Z UJ ::E UJ ...J l I UJ '" I 20 40 5 60 7 8 - ---~- -=--+.- '" ". I i "'- 1 I -" ! I I ~ I I -----+---- - ; , "-- I 10 100 1000 TIME (min.) FIGURE 7 TYPICAL CONSOLIDOMETER SETTLEMENT-LOG TIME PLOT 25 800 '-+---+---~---~N o l e( cr ~ ~---~---------~--~~~--~o ~-~-~---~--~-~--~--~~9 (f) EMBANKMENT Fig. 2: Geological Section ~ I > (f) G.l I i This method is generally accepted for sands but it's use for settlement prediction for clays is less well substantiated. Never theless it is believed that for the initial estimates the method is of considerable value. (Bache lier and Farez, 1965; Gielly, Lareal and Sanglerat, 1970; Jones, 1975). In addition ' to using the CPT results for settlement estimations, they are also used to give a preliminary indication of undrai ned shear strenghts (Cu) for soft clays. For this purpose, a conservative value of N = 20 is usually adopted in the following equation: 419 C u It should be noted that for normally con solidated clays in South Africa, the value for N found from comparisons of CPT and triaxial testing is generally between 15 and 20 but this may be significantly higher for overconsolidated clays. The boreholes and CPT's (Fig. 3 & 4) showed that the flood plain subsoil con sists of about 2,5 m of sand overlying 7,5 m of soft silty clay over about 3 m of sand over the shale bedrock. Preliminary analyses of the boreholes and CPT results indicated that the estimated settlement was about 0,7 m and the undrai ned shear strength approximately 10 - 15 kPa. A total strength analysis utilising stability charts (Pilot and Horeau, 1973) showed that the embankment would have a Factor of Safety of about 0,9 with a height of 5 m and side slopes at 1:2. 0,0 2,0 SATURATED COARSE SAND -- : BLACK SHATTERED WEATHERED - SHALE . L-_-L_---I L---L __ I--_-'-J 2 3 20 40 60 VOID RATIO eo WATER CONTENT 0/0 Fig. 3: Typical Borehole Log The nature of the clay suggested that the permeability might be low, possibly giving rise to problems with rates of dissipation of pore pressures, but it was noted that numerous sandy lenses were encountered. These results showed that further inves tigation was required, particularly since information regarding the permeability of the subsoil was essential. 3.2 Second phase investigation This consisted of seven boreholes for both 100 rom and ISO rnrn diameter piston samples and in-situ permeability testing. The smaller diameter samples were used mainly for visual fabric assessment, whereas the larger samples were used for laboratory testing. Conventional consolidation tests on SO mm diameter samples, and Rowe Cell tests on ISO mm diameter samples, with vertical and radial drainage, were carried out. In order to obtain samples for the latter it wws n~cessary to manufacture a ISO rnrn diameter thin wall piston sampler. In order to minimize disturbance, the tubes were kept fairly short, i.e. 500 mm. No problems were encountered in retaining the samples in the tubes during extraction from the boreholes, even when a very sandy lens was sampled. CONE RESISTANCE MPo - J 2 3 0r-----~--+-~~----~--------4 I .... Il. W a 10 , .... ) < FRICTION RATIO ------ 5% Fig. 4: Cone Penetration Test 15% It is frequently advocated that in-situ permeability testing is the most reliable method for assessing the time settlement characteristics of subsoil due to the advantages in testing effectively, very much larger and comparatively undisturbed samples (Lewis, Hurray and Symons, 1975). In-situ permeability tests were carried out by connecting a portable constant head permeability apparatus to twin hydraulic, porous ceramic piezometers which had been installed in a sand pocket vf controlled dimensions. The results of the in-situ permeability testing ar e shown as Cv's in Fig. 5, cal- culated from the field permeability and laboratory Hv's. 4 LABORATORY TESTING Typical results are given in the tables below and shown in Fig. 3, 5 and 6. Table 1. Atterberg limits, natural moisture and particle s i ze. Oepth (m) 2,75 3,90 4,50 5,00 6,50 1000 <; N~ E > u " 36 41 43 43 29 w % p 20 23 27 22 13 EI 0 X . w % % <0,075 mm % <0,002 mm 47 61 51 80 13 60 71 28 58 83 24 42 53 17 IN SITU PERMEAEILmES ROWE CELL (RADIAL) ROWE CELL (VERTICAL) 50mm CONSOLIOOMETER. .----. 50 100 150 200 250 300 350 400 cS' kPa Fig. 5: Cv versus Effective Stress Tabl e 2. Triaxial tests - Undrained and drained. Oepth (m) C kPa C' kPa 0" u 3,2 15 5,0 11 15 25 6,0 16 20 19 6,2 12 6,5 2S 420 Tab1 e 3. Conso1 idometer - 50 mm diameter. Depth (m) M m2/MN (l00 - 200 kPa) C mi /year v v 2,75 1,47 0,5 4,5 0,67 1,1 5,0 . 1,06 O,B 6,5 0,43 2,0 B,7 0,49 2,0 Table 4. Consqlidometer tests - 150 mm diameter ROle Cell vertical and radial drainage. Effect ive Stress kPa Depth (m) 14-23 23-56 56-112 112-224 224-392 4,00 C mZ /yr 220 45 2,0 1,0 0,7 Vertical MV mZ/MN 0,33 0,31 0,71 0,10 0,53 v 4,00 C 162 19 2,0 1,7 1,5 Radial MV 0,30 0,41 1,1 5 1,29 0,57 v 5, 50 C 31 14 12 4,0 3,0 Verti cal MV 0,41 0,41 0,71 0,71 o,n v 6, 50 C 51 24 3,0 Radial MV 0,60 0,45 0,59 0,53 v In the calculation of the coefficients of consolidation in the above table, the square root-time method was used for verti cal drainage with the Terzaghi time factor 0,848 for 907. consolidation, and for the radial case the 0,465 root method with a time factor of 0,334 (Silveira, 1953). In addition to the above tests, dynamic consolidation tests were carried out by Techniques Louis Menard in Paris since the Menard dynamic compaction technique was being considered as a possible field treat ment. These tests however, together with chemical analyses, indicated that the sub soil was not suitable for the Menard method. Geological history indicates that the material is a normally consolidated recent alluvium. The laboratory tests and the visual description of the dominant subsoil material show that it is highly compress ible, very soft to soft silty clay. Local experience has often shown these deposits to have a higher strength surface crust, overconsolidated due to dessication, which can be of significance primarily because it allows reasonable trafficability on the site. The higher strength is usually dis counted in stability analyses because of the possibility of tension cracks arising from prior shrinkage or settlement under the embankment. At this site however, there is no evidence of overconsolidation 421 in the clay, both fron the e-log p curves, from comparisons of Cc from these and from Liquid Limits, and from Cu measured and calculated from Plasticity Indices. It will be observed that Fig. 3 shows that the natural water contents, almost without exception, exceed the liquid limits, i.e. the average Liquidity Index = 1,54 which is somewhat higher than was expected. o o > 1,4 1,3 1,2 I, I 1,0 0,9 0, 8 Go-- SAMPLE 3,9 m ~ \ 1\ 1\ 10 50 100 200 400 EFFECTIVE STRESS ' cr' (kPo) Fig. 6: Typical e - log 0 1 5 ANALYSES The analyses are considred in two parts, i.e. settlement and stability. 5.1 Settlement The coefficients of volume change Mv , were fairly consistent both from the small dia meter samples and from the large diameter tests with vertical and radial drainage. On this basis, the estimated settlement was 0,8 m, which is similar to that predicted from the preliminary Cone Penetration Tests. Since it was clear that the rate of con struction would be restricted by the excess pore pressure dissipation rate, it was not possible to calculate the overall time for settlement until the construction rate had been determined. Simplified calculations however were carried out to establish the order of time, assuming instantaneous full height construction, using the Cv's from the Rowe Cell laboratory tests and those derived from the in-situ permeabilities. It will be seen from the typical results given in Table 4 that there is no signifi cant difference in the vertical and hori zontal coefficients of consolidation; be- cause of this, and the large embankment width, the problem was therefore considered to be one of plane strain with single ver tical drainage only. and equalled 807. of the load increment. These estimates indicated that the tine for consolidation, would be about 2,5 years to which should be added the construction time. A similar calculation using the Cv values from the small diameter consolido meterS, gave a time of about 10 years, which would not have been acceptable. The actual time for construction, calculated later as 3 months, has to be added to the 2,5 years, but it must be noted that the estimate is for 907. consolidation. The remainin~ consolidation settlement, say 80 mm, together with a similar allowance for secondary compression was considered to be acceptable for the embankment itself but would cause problems at the bridge due to negative skin friction. It was there fore decided to surcharge at the abutment position with an extra 2 m of fill. 5.2 Stability Stability analyses were carried out using , the Bishop Method of Slices assuming a circular failure surface. These showed that the maximum height which could be built, assuming no excess pore pressure dissipation was 2,5 m if a factor of safety of 1,2 is taken as the lowest permissible limit during construction. Similar calculations indicated that the long term factor of safety, assuming com plete excess pore pressure dissipation was F = 1,57. These calculations were made on the basis of strength parameters c' = 5 kPa and ~' = 35° for the fill which was shale from an adjacen~ cutting. The subsoil parameters were taken as c' = 0 and ~' = 25° although the laboratory tests showed a small cohesion, there was no evidence of overconsolidation, and this cohesion was conservatively ignored. The water table was assumed to be at surface and any bene fit from the sandy layer close to the sur face was ignored since this layer had been shown to be variable and practically non existent in some areas. The long term stability was therefore considered to be satisfactory so that the problem became simply one of determining the appropriate construction rate. The model used was the simple one of increasing the height of the embankment, above an intitial 2,5 m height, in incre ments of 0,5 m. It was assumed that the actual time to construct each layer was negligible and that the excess pore press ure response in the clay was instantaneous 422 The factor of safety was limited to 1,2 and therefore, at any height, an allowable excess pore pressure was calculated. It was then possible to calculate the amount of dissipation needed at each embankment height such that the allowable excess pore pressure for the next load increment would not be exceeded • . The times for dissipation for each height increment were calculated and summarised to give an estimate of con struction time. The purpose of this was to check that even with pessimistic assump tions, the actual construction rate would nevertheless be acceptable. Since in this case the time was calculated as 3 months this was considered as reasonable. This approach to the construction entailed a relatively high risk s ince the objective was to keep the factor of safety low and the rate of construction maximised so that the time available for subsequent settle ment could be maximised. The advantage of this was that artifici ally accelerated subsoil drainage or wide berms to increase the stability of the embankment were unncessary. An essential corollary of this overall approach however, is that the embankment must be monitored during construction to check the stability and to confirm that the actual field dissi pation rates are at least as high as the predicted values . 6 MONITORING Two lines of piezometers were put in, with three piezometers in each line as shown in Fig. 1. The piezometers were of the three lead, pneumatically operated bellows type with the leads taken to gauge houses out side the toe of the embankment. Readings were taken by connecting a portable com pressed CO, bottle to one of the leads. Stability control charts were drawn up as shown in Fig. 7 and these showed that the factor of safety was satisfactory. It would be imprudent to rely solely on pore pressure measurements to assess stab ility, and various strain measurements were considered to be an essential back up system. It is generally agreed that sett lement measurements alone are insufficient and that a combination of these and hori zontal deformations gives a better control. Settlement gauges and inclinometer casings were therefore installed in the positions shown in Fig. 1. The settlement sensors were of the same type as the piezometers, i.e. pneumatic, with the bellows pressures resulting from mercury filled leads to mercury reservoirs in the gauge houses. A total of six sen sors were placed approximately 0,5 m below ground level before the start of construc tion. 1,5 ..... ~I,4 .., ..... <{ (f) 1,3 ..... 0 II: 1,2 0 ~ u <{ 1,1 ..... 1,0 0 10 20 30 40 50 60 70 eo EXCESS PORE PRESSURE, Ue (kPc) Fig. 7: Stability Control Chart The gauge houses were levelled at the time of each reading, with reference to a nearb y bench mark installed into rock. The inclinometer casings were grouted into rock at their bottom ends so that absolute values of displacement could be measured. In addi tion to the settlement sensors descri bed above , simple water level U-tube sensors at different heights within the fill were also installed. In general the inclino meters showed very little horizontal move ments only averaging about 20 mm. However, one inclinometer, the eastern one on the northern line, showed a movement of 70 mm with the peak val ue at 2 m depth. It will be seen from the plan - Fig. I that this position is close to the loop of the river which was filled with sand af ter the diversion canal had been cut. The max imum displacement occurred at the old river bed level. As a precaution, the frequency of taking readings was increased; these showed that the rate of horizontal movement decreased rapidly as soon as loading stop ped suggesting that failure was not imminent. The ratio of horizontal strain at the toe, to vertical strain at the centre was plot ted in the manner described by Matsuo and Kawamura (1977) and shown in Fig. 8. It should be noted that even the one compara tively large horizontal displacement men tioned above does not indicate any problem on this chart. This would appear to be a useful method for describing the embankment behaviour since the latter can be seen qualitatively at a glance from the rate and direction at which the plotted line approached the factor of safety lines. It was not necessary to curtail the rate of construction due to any danger signs but this was probably because the rate of con struction was considerably slower than the maximum calculated since it was convenient to use the earthmoving equipment at an adjacent similar site. A typical plot of height of fill, pore pressure and settle ment is given in Fig. 9. The settlement is close to the predicted levels and the rates of pore pressure dissipation and settlement are in reasonable agreement with the predicted values . 3,0 2, co E "0 1,0 0+-~~-r~-4--r-~~~~ o 1,0 Fig. 8 : Hatsuo Control Chart 7 SUMMARY The investiga tion was carried out in pre liminary and detailed stages. The prelim inary work revealed the extent of the pro blems and allowed planning of an economical detailed second stage. Establishing realistic values of the sub soil permeability through in-situ testing and large diameter consolidometer (Rowe Cell) tests was a vital contribution to the overall design. The decision to build the embankment at a monitored controlled rate was based on the predicted behaviour and on comparisons of costs with other schemes such as insta llation of drains and utilising flatter slopes or stabilising berms. The monitoring has shown that the embank ment has behaved as predicted and has more than justified the cost of the extensive 423 - investigation through the minimizing of overall construction costs. 1979 APRIL MAY JUNE JULY AUGUST SEPT. OCT. NOV. -E - 4 ..J d "- :3 ~ ~ !i: 2 C> I iii J:: C 0' , "- 30 ~ UJ Q: :::> 20 (I) (I) UJ ~ Q: "- 10 UJ Q: 0 "- 1) 0 E E -200 I z UJ ~400 ~---- ---- PREDICTED SETTLEMENT. ~-­ -- ..J l I- --- ACTUAL SETTLEMENT. ~600~ ____________________________________ ~ Fig. 9: Typica l Honitoring Record 8 REFERENCES De Beer, E.E. and Martens, A. 1957, Method of computation of an upper limit for the heterogenity of sand layers on the sett lement of bridges. Proc. 4th Int. Conf. Soil Mech. and Fdn. Engng., Vol. 1 : 275 - 282. Bachelier, M. and PaEez, L. 1965, Contri bution to the study of soil compressib ility by means of a cone penetrometer. Proc. 6th Int. Conf. Soil Mech. Fdn. Engng., Montreal, Vol. 2: 3 - 7. Gielly, J., Lareal, P. and Sanglerat, G. 1970, Correlations between in-situ pene trometer tests and the compressibility characteristics of soilds. · Conf. on in-situ investigations in soils and rocks. London: 167 - 172. Jones, G.A. 1975, Deep Sounding - its value as a general investigation tech nique · with particular reference to fric tion ratios and their accurate determi nation. Proc. 6th Reg. Conf. Africa Soil Mech. Fdn. Engng., Durban, Vol. 1: 167 - 175. Pilot, G. and Moreau, M. 1973, La Stabilite des remblais sur sols mous. Paris, Eyrolles. Lewis, W.A., Murray, R.T. and Symons, I.F. 1975, Settlement and stability of embank- 424 ments constructed on soft alluvial soils. Proc. Instn. Civ. Engrs. Part ~ 59. Dec. : 571 - 593 . . Silveira, I. 1953, Consolidation of a cylindrical clay sample with external radial flow of water. Proc. 3rd Int. Conf. Soil Mech. Fdn. Engng., Swi tz. , Vol. 1 : 55 - 56. Matsuo, }!. and Kawamura, K. 1977, Diagram for construction control of embankment on soft ground. Soils and Foundations Japanese Soc. Soil Mech. Fdn. Engng., Vol. 17, No.3: 37 - 52. 7/19 200 400 600 'J kPa SPT 10 20 30 qc o Or-------~========~--~~ MPa N 2 4 18 20 (c) 0 2 4 6 ( Uo Fig. 4 ~~ --. , " ,. .' '. UI Silly land hy~aullc fill Black sof! clay I sand lenses Grey silly sand I dense to very dense Sofl silly clay Typical result under storage shed 26 10 12 5 4 III 14 III III III 16 III III 33 III III 12 I I 3 Gypsum tailings dam. An existing and practically complete dam was tested to obtain data on the state of consolidation since the possibility of reclaiming the dam was being considered. A typical probe result with borepole log is shown in Figure 5. 100 200 300 U kPa 2 3 q cMPa Gypsum ~ I' Slimes ~-~> ~ ,~ ~ c:.::"_ ~ (". -.:-: r-~=, - ~:.:~. ---- UI Very ---- lofl \ -~~ ......... cloy \Uo _So Fig. 5 Typical result through gypsUm tailings dam 492 (d) x ~ ... Q o 00 01 12 Gold mine tailings. A dam was probed, primarily to establish the layering and relative permeability of these layers, so that the most appropriate modell ing of the dam could be made for the purpose of stability analyses. Typical results are shown in Figure 6. No borehole log was available for the subsoil, nor was this penetrated by the tests. .... " .... ~:.="'--- 200 1 :::-----"::.~~ =~.:-- - - - - - - - --..-----=- -=- - - ~. " - - ----..- Fig. 6 Typi cal result from gold mine tailings dam (e) Platinum tai l ings . A dam was probed for the same reasons as t he gold tailings dam and typical results are shown in Figure 7. Information on the subsoil was available and this consisted of stiff, black, shattered and slickensided slightly sandy, silty clay derived from residual decomposed .norite with WL = 65%; Ip ~ 38% and a clay (0 . 002 ~m) fraction of 60%. DISCUSSION (i) Consolidation. It is emphasised that the primary purpose of the test programme was to establish semi empirical correlation.s of field and laboratory consolidation data. Many field dissipation tests were carried out and facsimiles of parts of typical results are shown. in Fig. 8. The upper part of the figure shows the pore pressure response during penetration, i.e. the chart is on a depth base. The lower part shows dissipation tests of pore pressure against time. Superficially the results are similar to laboratory consolidation test data and a similar processing method is therefore adopted. o 10 " 12 13 Fig . 7 0 E 2 J: >- Cl. 4 '" c 6 5 " \ \ 100 \ • kh 4e UP. zoo z ---~ -- - -- - ---::.~- --:.-"S,~. \ '-L~~~c~r_-_-_-_-_-_-_-_-_--_-_-_-' u. \ -Uj--- 'c , Typical result from platinum mine tailings dam PORE PRESSURE k Po 50 100 150 200 250 " so. FIG 3 se. FIG 2 "- '.:: "- Uo "- Stort of Fig. 8 Typical dissipation test results 7/19 The right hand part shows results taken from the probe test shown in Figure 2. It will be seen that on stopping penetration there was a practi cally instantaneous decrease in pore pressure and that this portion was sometimes a significant part of the total dissipation. This was seen very clearly on the actual charts of test results and it was necessary to switch the chart recorder to the time control immediately before stopping penetra tion to correctly record this. It was tentatively suggested that this immediate decrease was a dynamic effect due to the stopping of penetration and should therefore not be considered as part of the dissipation. A data correction procedure was therefore adopted of plotting the results on a square root of time basis as shown in Figure 9. Fig. 9 Typical pore pressure against square root time A practically straight line part to the curve resulted, which intersects the pore pressure axis. This point was taken as the dissipation pore pressure datum, u eo ' with the time axis unchanged and the datum as to' Final dissipation was taken as tlOO, IJe100 ' and the parameter for the subsoil consolidation characteristics, tso, was defined as the time at which half the dissipation had occurred ueso . The tso's were read off the pore pressure against square root of time plots. The left hand side of Figure 8 shows the result from the probe shown in Figure 3. On stopping penetration the pore pressure increased before subsequently decreasing. The reason for this was not established but it wa~ presumed to be partial blocking of the filters which in this case were the face mounted elements possibly with some loss of saturation. It was noted that if the results were plotted on the same square root of time basis as the previous case, then a straight line section intersects the time axis at a real time. 493 7/19 494 This was taken as the time datum, to' by shifting the time axis; the pore .pressure datum, ueo ' was taken as the maximum pore pressure recorded. Because of the possible doubt concerning the valid ity of these results they were not used for corre lations with laboratory data, it was however inter esting to observe that the slopes of the straight line portions, and the tso's deduced, appeared to be in close agreement with those derived from the much more usual dissipation tests in which immediate decrease of pore pressures occurred on stopping penetration. It is possible therefore, that although partial blocking of the filter elements inhibits response during penetration, it may not be significant for the much longer times for dissipation tests. In the type of tests described above, the time for full dissipation may be long and inconvenient, an alternative method of arriving at tso, or an equivalent parameter, was desirable. If the tests were being carried out at a site where the in-situ pore pressure regime was simply the hydrostatic head, then if this was known from the water table, the u ,can be taken from the root eso time plot without the full dissipation being achieved, provided that it was continued beyond the tso time. This made a considerable difference to the time required for testing and made the procedur~ very much more convenient for use as a routine field test. If the pore pressure regime was not known, it was considered possible that a deduction of the permea bility could be made from the slope of the straight line part of the dissipation results, (see Figure 9). It was found however, that although the slopes, which are the initial rates of dissipation, gave a relative indication, it seemed to be too insensitive to use as a quantitative measure. It also implied a square root time modelling which may not be valid · and was therefore not pursued further. Typical tso's from the results of dissipation tests carried out at a site at which comprehensive laboratory large diameter (150 mm) . as well as stan dard 50 mm diameter, consolidation test data was available, are shown plotted against the reciprocal of the coefficient of consolidation in Figure 10. A best fit straight line through the origin was drawn and resulted in the following equation tso ( . ) 50 m1ns = - (2/ ) C v m year It is suggested that this equation should be seen as a pre liminary attemp t to es tab lish a useful field relationship and that much more data is needed to refine this or similar correlations . The ease and economy of cone pore pressure testing (CUPT) should be borne in mind so that the advan tage of being able to obtain consolidation data from simple fi~ld testing is fully exploited. It may also be noted that considerable scatter usually occurs in laboratory test consolidation data resul ting in difficulty in selecting appropriate values for analyses. A greater number of field tests may provide a better statistical selection procedure. It was incidentally noted that the field tso's were generally about an order of magnitude higher than laboratory tso's where the consolidation test equipment had a drainage path length of 10 mm. o " . 90 0,50 Yev year/m 2 1,0 1,5 Fig. 10 Correlation of field dissipation times with laboratory consolidation data (ii) (iii) This implied that the effective length of the field drainage path was about 35 mm which intuitively seemed reasonable for a 35 mm diameter cone. In-situ pore pressure measurements. A secondary purpose of the field testing was to measure excess pore pressures under an embankment which was being constructed. A number of the piezo meters had shown that high pore pressures existed and that although settlement appeared to be as expected, the pore pressures showed very little dissipation. Holes were drilled through the embankment and CUPT 's carried out into the sub soil. These showed high pore pressures under the embankment with the values in agreement with the piezometers. They also showed that the tso's were higher for material under the embankment than for material at the same depth outside the embankmer.t which was expected, since the laboratory work (Jones and Rust, 1981) had shown C v to be highly stress dependent. Subsoil Identification. All the results of CUPT's, where independent information on the nature of the subsoil was available, confirmed observations by authors referred to previously that the pore pressure response is extremely sensitive to the nature of the subsoil. Sufficient data has not been analysed to develop quantitative relationships but if the suggestion by Baligh et al (1980) was adopted of using u/ qc as an identification parameter, then the results from natural materials shown in Figures 2 and 4 indicated that for soft clays u/qc varies from about 0.4 to 1.0 which agrees with the Baligh et al results. Loose sands showed u/qc values of about 0.2 to 0.5 and dense sands have lower values. The gold mine tailings dam u/qc values were extremely variable (as shown in Figure 6) from about 0.02 where the cone pressures were highest, to about 0.3 for the lower cone pressure zones. These were assumed to reflect coarser and finer layers respectively, resulting from the method of construction of the tailings dam. The platinum tailings dam results (see Figure 7) showed similar results except that the variation was considerably smaller; both in the thickness and number of layers and in the u/q value which c were typically from 0.02 to 0.10. The results from both these dams showed remarkably good definition of layers indicated by the virtually totally consistent resu1~s of low cone pressures accompanied by high pore pr.essures and vice versa. Figures 6 and 7, because of the size of the drawings, are much less detailed than the field chart results where the reversals of cone and pore pressures were strikingly obvious. The charts enabled layers of about 25 mm thickness and less, to be clearly distinguished. The gypsum slimes dam generally showed less pronounced layering with an overall u/qc of greater than 0.5 The layering in these dams, which varied spatially and in consistency, undoubtedly has a most significant influence on the overall permeability. This may be taken into account in the development of the most appropriate flow nets for dam design purposes and could also affect construction practice. However, despite the apparent confirmation of the use of u/qc as a material identification index it is suggested that some caution is required in that in many cases the selection of representative values from the fie 1d resu1 ts 1n a multi layered subsoil system is difficult because of the rapidly changing values of both pore pressure and cone pressure. A further potential difficulty is that in some cases (see Figure 2) the pore pressures, in a layer which appeared to be very consistent, increased much more rapidly than the cone pressures so that there was no unique u/qc value. This suggested that a measure of the rate of increase of pore pressure would be a more approp riate parameter, which would appear to contradict Ba1igh et a1's findings. It should be noted how ever, that the rapid increases in pore pressure with depth, referred to above, occurred between stoppages for dissipation tests; a method of correction for these was not defined, nor was the overall effect of the repeated stoppages clear. Because the emphasis of the work described here was on dissipation tests from excess pore pressures, ue ' down to hydrostatic pore pressures, uo ' the authors suggest that -ue may be a mare sensitive measure than ut ' for defining a material identification parameter; this could lead to a normalized subsoil index, I , defined as follows : s where I s (u t _ uo) IUo (qc - avo) lovo vertical total stress. 7/19 Insufficient data has been processed for a thorough comparison of u/qc values with the above. The values would obviously be different and in many cases would be negative. This could possibly be an advantage in that it would lead to a clear indication of -di1atant materials. However, it may be noted that the expression given above is effectively a measure of the pore pressure parameter at failure, Af , and it seems unlikely that this would usefully define the nature of the soil but would define the state of consolidation. An alternative materials index may perhaps be simply the ratio of excess pore pressure, ~e' to the hydrostatic pore pressure, ~o' i.e. (~t - ~0)/~0' CONCLUSIONS The use of a pore pressure measuring sensor in a probe to give a cone with pore pressure penetration test (CUPT), considerably enhances the data obtained from probing. At present there is insufficient theoretical basis for the full interpretation of the pore pressure information obtained . It is therefore suggested that efforts should be made to establish semi empirical correlations. In-situ pore pressure dissipation tests were carried out and the results compared with laboratory data. These led to the following relationship t50 (mins) The pore pressure probe was also used for measuring pore pressures under embankments during construction and gave results in agreement with in-situ piezometers. Data from the CUPT's was also used to confirm the use of u/qc as a materials identification parameter. Although the results are encouraging, it is suggested that this parameter may be more suitably defined in terms of excess pore pressures rather than total pore pressure. ACKNOWLEDGEMENTS Gratitude is expressed to C. de Bruin and E. Rust for assisting in the development and manufacturing of the probe, and to the latter and K. Anderson for conducting the field work. The encouragement and co-operation of the South African Department of Transport is also greatly appreciated. REFERENCES Ba1igh, M., Vivatrat, V. and Ladd, C. (1980). Cone penetration in soil profi-ling. ASCE, J. Geotech. Div. (106), GT4, 447-461. Begemann, H. (1965). The friction jacket cone as an aid in determining the soil profile. Proc. 6th ICSMFE (1/4), 17-20. Montreal. Janbu, N. and Senneset, K. (1974). Effective stress interpretation of in-situ penetration tests. Proc. ESOPT 181-193. 495 7/19 Jones, G.A., Rust, E. and Tluczek, H.J. (1980) . Design and monitoring of an embankment on alluvium. Proc. 7th Reg. Conf. Africa SMFE, (1), Accra. Jones, G.A. and Rust, E. (1981). Design and monitoring of an embankment on soft slluvium. Submitted to 10th ICSMFE, Stockholm. Parez, L., Bachelier, M. and Sechet, B. (1976). Pression interstitielle deve10ppee au foncage des penetrometres. Proc. 6th Reg. Conf. Europe SMFE, (1,2), 533-538, Vienna. Schmertmann, J.H. (1974). Penetration pore pressure effects on quasi static cone bearing qc. Proc . ESOPT (2/2), 345-351. Schmertmann, J.H. (1974). General discussion, pore pressures that produce non conservative qc data. Proc. ESOPT 146-150 . Torstensson, B. A. (1975). Pore pressure sounding instrument . Proc. ASCE Speci ality Conf . on In situ Measurement of Soil Properties, (2), 48-54, Raleigh. Wissa, A., Martin, R. and Garlanger, J. (1975). The 496 piezometer probe. Proc. ASCE Speciality Conf. on In situ Measurement of Soil Properties, (1), 536-545, Raleigh. (vi) DESIGN AND MONITORlNG OF AN EMBANKMENT ON SOFT ALLUVIUM C. A. ';O' u 10 5 o 50 100 150 200 ~O 300 350 400 Fig.3 ANALYSIS O"kPa Coefficients of consolidation versus effective stress The stability and settlement analyses were carried out :following each of the two phases of the investigation. First Phase Stability Total stress analyses were carried out using the Bishop Method of Slices an~ also from stability charts, (Pilot and Moreau, 1973) w1th shear strengths derived from Cone Penetration Testing and from quick undrained triaxial tests. These showed that the embankment height was limited to 3 m for a factor of safety of 1.2. !The mean cone reading, qc' was 392 kPa, the standard deviation was 43.9 kPa and the coefficient of variation, lV~cwas · l1%. ______ _ . _________ . G.A. JONES 'Van Niekerk, Kleyn & Edwards, Silverton, South Africa E. RUST Van Niekerk, Kleyn & Edwards, Silverton, South Africa Tr l ' j·: T in.1-: 1II',k l" jJ ~ rJ kST I' .\(: j·: F: ·: !t. h • .IIId Fr t·'III.) 3 f.CR I\T.Z l e l loE T1T i{ E .\ 1.;\ ~1. \ Cl ll :--; I :. "i1}R LA PI:l : "" Ej~ r " ,\(:E \ En h , 11 \.1, .. "t '!J . \n ': !.I1 "') .~ .. _. __ . .. __ ._-_.-.- r~~~-_undrained - shearsi:reu"gth~w';i -~~iculated from "(Lunne, 12ide" ,and ,de "Rui ter, .. 19.7.7),., '" 1'" ... <: ~.,. ~ <:1 i I where q -c N -\r. cone factor I I' "[ - - f undrained shear strength Y - soil ooi t weigh t z - depth From the above, C v commonly accepted jidated soft clay. - 20 kPa if N\r. - IS, which is a value for this tlPe of normally consol- ffitJCt~~o,r~~ptY.:J~:s ,t~s"~ showed lassumed . to· be due to ' samp Ie dis turbance~ · iFirst Phase Settlement !Settlement calculations were based on the CPT results, iusing the Buisman - de Beer approach, (de Beer and ;Martens, 1957), but modified as shown below, (Bachelier. land Parez, 1965; Gielly Larel and Sanglerat, 1970; Jones, 11975 ). I I I . h h ' " . :-; , 1 ! i l": .IS III l i e S:lmp c cne CS CU WIt. ' C u~ .. t ·t!n .. oJ . - I 'i!'ricdy wit hin the liln iuo;it iG;!cr!. 2 r~'3 H··· ·l~ ·:'.is ~ :('!' z '~ -t: ocr) I ~ap- 1,4 I- '" IL <;1- - .0_ - . ;:~ · - - I-~ · - '- _ . ' / l-I_ - / ~ I -- - ' 17 I : ~~ 1'7 1- ;7tc 1--, -f-f-- -:'7 L7 7" - f-- -- - 7' ~ .. / ~ - // o / IL lk L . 0 __ - "'" 0"'" tIlt 0'1 C\.A' Figure 3. Grading Envelope for Gold Tailings ... x ~ '" ~ ~ 15 z ;;: '" ., '" .. oc ... .., u.s. Sr,lNOAIIO Sl~ Sill 'I, u'"' ~'" ~ -:' / ~rX 'W r/ - J K2 I . J - _I~ I +-~ - - :- - - - 1- -- - - 1- . 0 -i-- · - - - :x- - - r- - -r- - - .,! ~ J - - - , '" I -IL -- . ~ .. / ., / ~., II -,- / / L [.L~ I . . _'-< ~ 0 potJ "" '"GRAlH SIZE IN ~~ .. .... """ COltlU lilt 0" ClA' Figure 4. Grading Envelope for Platinum Tailing~. i I I I· I I MINE TAILINGS CllARACTERIZATION JU7 This p"per will di"cuss the test results obtained from typical te s ts ill goid ",,,I platinum tai1JlIgs. It is shown how these results ;'Ire I llterpreted alld the cOllcen,s with such interpretation!!. !yp'ical Test Results Portions of the pore pressure ;'Ind cone pressure profiles obtained rrom the testing of gold and platinum are presented in Figures 5 and 6. The followIng general observations can be made from these results. (i) There is a marked relationship between cone pressure lind pore pressure response, L e. higher point resistances and lower pore pressures occur at the Bame depth while lower point re sistnnces are obaerved with higher pore pressures (e.g. 16 to 16.5 m, Figure 5). Th10 is assumed to indicate the lnyering of the materinl since, during penetration, the denser coarse materiai tends to dllate, resulting in II de crease in pore pressure, wherea!! a positive pore pressure is Cllused in the less dense finer material. It mAy be noted thnt lllyer thicknesses in Figure 6 average about O.25m but individual layers of about O.05m and less can be readily detected. (ii) Co'"p1ete saturlltion of the pore pressure system is of con cern in interpreting piezometer cone datil. It is di[[icult to identify a 'soft' system from testing results in variable material suci. as tailings. There may be some indications in Figures 5 and 6 that the cone was not fully saturated; note slow response between 17.5 and l8.0m in Figure 5, as well as the apparent smaller lind slow response of pore pressure in Figure 6. Extrapolating the hydrostatis pres sures in Figures 5 and 6, water table depths of about 6.4m and 3.2m are obtained ~espectively. These depths were there fore penetrated before ony positive pore pres!!ures were measured. It is possible that Borne desaturation could take place. The correct combination of porous element air entrainment value and cOile fluid, such as glycerill, controls the saturation of the cone when penetrating Wlsaturated materials. From other testing performed on tailings by the authors, the results in Figures 5 and 6 are considered rep resentative of testing in tailings under fully saturated conditions • . The further interpretation of the data will therefore assume saturated conditions. (iii) During probing it is necessary to stop after each meter ' to add a further rod. Thls stop tokes about one mInute and during the period, the generated excess pore pressure almost completely dIssipates. These pP!1i.tlons are indicated by horizontal lines e.g. at lS.5m and l7.5m in Figure 5. At these, or at other positions, complete dissipation ls some tinles allowed so that the. in situ pore pressure can be measured. J08 15 n . w 0: .. ;11 ! o r CONE PENETRATION 250 '00 no 1000 <=~~ --::.~~;=~- Dlltipalloft :::0 -:> ----- OI"lpotlo" ~e ~r. 12.50 U. 8 U, )Cr. ~e CONE pnESSUA[ V, PonE '"[5$V"[ V. HYO"OSTATIC '"[$$V"[ "I ~~~- .... .... D.pth Cone Hydr Total ~ota1. SoU Type + I +' Ret. (.) Pre •• ure Pressur'. Pore Pc. Vert.'r. o 0 Dcn~lty qe U Ut 0 (de~re •• ) D 0 vo r (llFa) (Ua) (kFa) (ltra) (X) (1) (2) Ol C4 1 - (5) (6) (7) (8) (9) 14.0 600 78 330 211 Cloy H 7J 0 U.O 6,100 111 100 227 Sand )) 3/, 1,7 U.II 4,400 9J 150 240 Silty Send 30 )) 34 16.4 600 99 650 250 Cley 13 25 0 16.6 6,000 100 180 253 Silty Sind J2 35 1,4 U.' 1,000 101 450 %S8 Cley 17 26 0 11.1 6,100 106 156 262 Silty Send .13 34 4S Figure 5. Ple~ometer Cone Rp~ul~R for Gold Tntllngn. MINE TAILINGS CHARACTERIZATION 309 o 6.0 · .. w or: 0- w ::t ! 1,0 Z'O , '00 no I I % 0- &. W o "-= ~ 0".1,011." '-.., (- ....... ., <' c::--.) -----. 9.0 ·~b ____ S Depth Cone (m) Pre88ure 'Ie (ltFa) (11 (21 5.4 3.161 5.7 667 S.' 1,000 6.2 1,500 . 6.5 5,000 7.2 667 7.4 833 1.5 3,100 8.3 333 lIydr Pres,ute U 0 (ltPa) . (3) 27.7 25.7 27.7 30.7 33.7 40.7 42.1 43.1 51.1 Totl1 :total Pote Pt. Vert.Pr. Ut " vo (kFa) (llh) (4) (5) , 21 7' 43 114 )) 117 67 92 21 97 200 109 116 112 )0 114 283 128 1 -- 'c .... 1000 IZ'" --- u. a VI k', I I 'c COH[ P"ts$V"[ V, PO"[ '"[$$U"[ U. HTO"OSTA Tit '"[$$V"[ Soll Typo +01 +~ Rol. I Dendty (d.ltU.) D r (%) (61 Icn (8) 19l S.nd 14 14 46 Clay.y Sand 21 24 0 Sind 11 3S 41 Clay.y Sand 29 32 19 Sand ]7 36 59 SlIty Clay 11 3D 0 Sllty Clay 2. )0 0 . Sand )) )) 41 I Clay 14 25 0 Flr,lIre 6. rtc~ompter Cone Rp,nllt" for Platinum Tnilings. )10 CONE PENETRATION Interpretation of Test Results Figures 5 and 6 show interpretations of the "ie7.om~t('r cOile test results at selected depths. The first five colllllln9 in the tnt.!.!.'!; r. 1ve depth from surface,cone pressure (qc)' hydrostntlc pre","lIre (II , ), tl'ta!. mensured pore .pressure (u t ) lind total overburden pres!lure ("vo~ l;,klnB dry unit weight at 13 3 kN/m? (85 lb/ft3) and saturnted Ulllt welBht ;,t 16.5 kN/m 3 (105 lb/ft j ). The methods used for interpreting the results nre discussed in this section. Also discussed nre the conce[1W wlth these interpretations. (i) t~terial Identification Various methods have been proposed for the idcntlficntlon of materials from cone penetration test results. negemnn" (1965) intro duced a semi-empirical method based on fdctiou ratios wldch hns bccn sho~m to be a very reliable system. lIowever, because of the sIO,,1l thickness of the layering in tailings impoundments (us ually less than 50 mm), friction ratio is not a practical way of fdentiCyinB tailings materials since the friction sleeve is itself 13[, nl111 loug. Baligh, et 01. (1980) suggested the usc of a u 1'1 rntio 3S n materials identification parameter and showed thnt ~ucr; a trltio Is useful in the identificat-ion of clay layers with differcnt over consolidation ratios. It was also concluded thnt II 1'1 decreases with decreased cone angle of tim penetrometer. Jones an5 V~II Zyl (l911l) suggested that indices based on the excess pore pressure (lie) may be a more sensitive measure. Note that u t a U o + ue ' where lie can be positive or negative. Three different soil indices ..,111 therefore be exmull1ed for the identification of materials: lsI - ut/qc Is2 • u - u t 0 u 0 Is3 - (u - u )/u too (qc - °vo)/°vo (1) (2) (3) In orde r to compare these indices, they are plot ted agnlnst cune pressure in figure 7 to 9. The data points in these figures nre frol1l test results in gold and platinum tailings below the water tnble. It can be seen that Figures 7 to 9 are all s1milar- and take the form of fairly narrow hyperbolic bands for the tallings materinl test ed, and there would appear to be no marked advantage in ony one of the plotted indice9. As previously mentioned, it is believed that the use of u e ' which may be positive or negative, has the acivantnge of indicat ing materials which are dilatant, but hos the disndvnntoge thot it is nccessory to know the naturn1 hydrostntlc cOIl-- SILTY SAND "' 0 00 pi ~- • --.. ..--------;; -. ~ ---= flJ _ ' 00 - SAND "--.. . " . --- --- - " --'" pi -.1!../- ~ , ~ 31) laond .IIL ____________________________________________ ~ o 10 20 30 ... 0 !lD Fir.llre Ip. qc-(ho OYO 5011 Identification from Piezometer Cone ResultA for ltorl1l :d . l)' C,,,,,,,,II":olr' " !;alll':llc,1 Malerlal .lJ4 CONE PENETRATION GCllerally it is confirmation tlHlt higher cone resistnnccs togethcr with smaller pore pressure responses are expected JII sallds ·, as cOnlp:lred to soft, nomally consolidated clnys, where small cone pressures 11nrl h1r,h pore pressure responses would be anticipated. The fairly nnrrow [,nlld of results, with comparatively sm~ll overlnp of dlrferent mnterlals within the bond, suggests that this plot can ""efully be deve toped n" an indication of the effective groin size of the IIInteria L. It is tentatively suggested that Figure 10 C:JII be IIsed In the form presented to identify normally consolidated saturated mnterinl" frolll testing with a piezometer cone having a 35.7 mrn, 600 tip. Snlld mnterials are larger than the no. 200 sieve, while clny i" sm"llcr th"n 0.002 mm (refer to Figures 3 and t,). It must he clllphn"lzed thnt the divisions shown in Figure 10 are tentntive, The!'e dlvl"lolls C:III he refined by well controlled laboratory testing. The absolute values of the parameters in Figllre 10 nre "trongly dependent on the measured total pore pressure and COliC reslRtnncc, so that it is a function of the shape of the cone, the po s J tiOIl of the filter, and the rate of penetration. Figure 10 was used to interpret the data ill FJr,lIre" 5 nnd fi, the resulting soil types are givell in ColulTUl 6. lis expected, cOllsldernhl e variation of soil types is obtained. (ii) Pore Pressure Heasurement Figures 5 and 6 demonstrate thnt the cone pre"sllre 11l1d pore pressure rendings change very rnpidly • . 11 careful nlln]y"b; o( the rl"ld recorded charts indicates a slight offset of th" pe:1k" nlld trouGh" (In cone pressure alld pore pressure rendillr,s). The pore premllOr.c response seems to lag approximately 40 nom beh.lnd the COliC preSRllre reading. The center of the ring filter element is about /00 IIIIn loJr.loer than the cone tip (refer to Figure 1) so that, I)uperfic.lnl.ly at len"t, it would appear that the · response lag is reasonable. The Incl .. ",Ion o( "', electronic data logger system sllould. help in the InventJgatloll of this effect. It shoul.d be . possible to ..inVl!stigate this problem hy laboratory testing of a carefully prepar~d layered nnn~le. There has been considerabl~ speculation regnrdtllr, the opt 111110111 p"nition oC th" ftlter clement. Torstensn,," (J975) drllloll,;tr:otrd th"t this has a marked effect on the magnitude of the pore pre"slIre IIIr,w ured during penetration. Parez, et al., (1976), however, obs"rved 110 such differences. 1I0wever, these results were comparing rel.ntively close poaitioning oC the filters. Jones and von Zyl (1901) nlso [o\lIld no differences with similar close filter positioning nnd experiellced frequent blocking of the face mounted filters in soCt clnys, which made the Cace position impractical. Levadoux and Daligh (1980), present a theoretical ann lysis of pore pressure generated during penetration. Figures 11(0) nnd l1(b) show the theoretical pore pressure distribution (normalized with rellpect to e f Cective overburden pressure) and the normalized pore p ressllre vn rin tion along the tip of a 600 cone. The results in Figure ll{b) indlcnte th;'lt the maximum excess pore pressure occurs olong the Up of the cOile. , I ."\ ;,~ 1~· If 1;j .:i: "r. ./t ., i' ' ! ~ .';1 MINE TAILINGS CIiAIlACfERIZIITION 315 From a "racticnl point of view, it is better to meo~ure the pore pres s ure just nhove the til' because of problems with smear and damage of fllter elelllents • --I 6U r;- Hr,l.Ir e 11. 00· !I 3 UUl.,. AT ,., ABOVE n, .~.- o 0.' 10 I . ~ 2.0 2.5 >J) AU!tAUJ,JII (a) (b) Theoreticnl rore Pre~!3l1re Ilistribotion for Probe Testing (60" ~one)(from Levadoux alld Daligh. 1980) It is essential thnt sonIC stondardized form oC probe be developed hefore n lIIeoni11g£l.Il comparison of moterinl identification indices, bosed on r,eneroted pore preooures, can . take place. It is suggested thnt the design shown in FIgure 1· is the· most . suitable for practical usc bnsed on the previously stated reasons. It olso conforms to the recoomnended shope for 0 Dutch cone given in the lSSHfE standard (1977). ". !.I££l.!'.'~~oent oL.!!y.Erost'!.!:.!E-Sondltlons 111 the Investigation of tailings impoundments it is necessary but often difficult, to establish the pore pressure condi tions which influence the stability. This is due to the cOlllplex lnyering system which may result in numerous perched w" te r tnbles. Furthe tmo re, flow is nea r-horizon tal in the pool nren nlld becomes drawn down at the embankment (see Figure 2). Nany fixed piezometers would be required to monitor these conditions completely. and, because of the thin lnyers, it would be necessary to ensure that any piez ometer only measure the pressure within one layer. In practice this is very dif f!cult· to achleve. .116 CONE PENETRATION IIowever, the piezometer cone can measure the pore pres"lIre at any point by simply stopping penetration and aUow!n!; dis sipation, even within thin layers since the fUter eiement has been deliberately reduced to a width of 5 nun. Figure 12 shows pore pressure measurements taken in this manner at the edge and at the center of tnilin~s iml'uul1dlllcl1Ul wi tho the water tables (measured with stnnd,,! "e l'ie7.olllcte n;) superimposed for direct comparison. It cnn be seen ~.at nt the center the measured pore pressures In meters of water (shown dashed) increases in exact corrcspondence with in creased depth which conforms to uniformly hor h:ontal or 7.CtO f low conditions. Close to the embankment, however, the m"ns ured pore pressures increase with depth at a lower rnte - indicated by a solid Hne in Figure 12. It is believed thnt this is a reflection of the changed flow conditions lit the embankment which can be shown by an """ropriate flow IIet (refer to Figure 2). Since it is the pore pressure conditions close to the cmh:lllk ment that control the stability of the dam, it is helleved that the piezometer cone will be invaluable in the assess ment of e.xisUng impoundment stability . b. Pore Pressure Dissipation Following Penetration Typical curves of pore pressure dissipation In soft clays following penetration were presented by Jones nnd von Zyl (1981). Figures 5 and 6 indicate thnt pore pressure di s 51."n tion tokes place each time a rod is ",hied (c.g. nt tS.5 III and 17.5 m in Figure 5). In the testing of cloys it has been observed thnt the exc::wr -~ - 10 I...!....!--lJ:J: I'-Y: F.J :i±L!-!-- I=tH.+ -.!-!-+P..y..;±I:..J.+j=t-f+!-=f:~;.++J. . H-'+H+' ~l-It:i:ru=!~ml: o - > b ....... U .,. Figure 13. 5 1 15 ~ . '-, 25 35 ~5 ( +) (degrecs) Graphical Repr~§~ntation of Equation 4. 319 , (iv) Relotive Density After extensive eophisticoted loborotory teBting, Schmert m""" (1970) fOllnd an empirical rciotionship between cone pressure (q ), vertical effective stresB (0' ). nnd relative density (0 ), which c "opplies to nonnolly consolldXYed uncemented, primorily q5ortz, geologically rec21lt, gaturated, f,ine SP sands in ,situ, "hen tlsing the Fugrotype. 10 cm , 60 , cylindrical ~lp, jdvonced continuously at 2 c,./sec. II For 'Ie and' 0v~ in kgr/cm , (= 100 kPo): D r. ~ 100.l~ ('1 /12.Jl 0' 0.71) (7) r 2.91 c vo The values of relative density calculated from equation 7 are given in collJlml 9 of Figures 5 and 6. Negative values were obtained where Dr ~ 0 is given. The relative density values are in general very low and compare with other results obtained by the authors in tailings impoundments from Dutch cone and piezometer cone testing. This is of concern because the relative density is a very important pnramcter in ~ssessing the seismic stability of tailings impoundments: This specific area requires further work before definitive conclusions can be reached about relative density values obtained from equation 7. Thc Piczometer Cone as a Honitoring Instrument The piezometer cone can bc used rather .1neltpensively to obtain geotechnical parameters from in situ testi~g. Interpretation of the teRt results can be done by using the methods described , in this paper. Undoubtedly these methods will be refined in the future. 32U CONE PENETRATION The construction of tailings impoundments contllllles durine the mIne's operational life. There are, therefore, opportlOnltl.ea to change construction procedures, if necessary, to improve the stnh.llity. The . use of relatively inexpensive in situ tenting procedurer., slOch ns piezometer cone testing, holds cOllsidernble promise in the IlIt'nitur.lng and design of such improvements. Piezometer cone testing results can be used ns the b.1sis for stability analyses of tailings impoundments. Careful OIOIII.tOrillg or pore pressure and movementa of the impoundment should [ollow such . stability investigations. Piezometer cone testing can then he repeated on on annual (or longer interval) basis to rl'-ev:11 u.,te the stability as well as the effects of changes in construction procedures. Such monitoring will help in identifying possible trouhle fipots and will [orm a good data base for the understanding o[ the behavior of the tailings impolnldment. It is furthermore possible that the mine can own p.lezometer cnne e'luipment for intermittent testing. Such equipment can be mounted on the back of a pick-up to ensure easy access. The piezomcter cone testing holes can also be used for the installation of standplpe piezometers. Conclusions TI,e piezometer cone is a very ufieful geoteclmlcal site lnveGtlgn tion tool which is now sufficiently developed (or gener.11 field investigations. The piezometer cone can be. used inexpcnslvely in mille tailings impoundments to: (i) Identify the materials in the tnlllng'J impoundment, e"l'eci".lly the layering. Considerable knowledge obout the sp:1tial variation of material chnracteristic!! cnn therefore he obtained rather inexpensively. (ii) Estimate the effective streneth pnnllnctem lind relative density o~ the t.a~lil!gs mat~dal. (iii) Estimate the consolIdation parameter!! through pore preSfiure dissipation tests. Because piezometer probe testing can he co",lucted .llIexl'ell",1vcly (1n relation to undisturbed sampling, if such snmplillg I,. possible at all), it can become a useful monitoring tool for minI? taU Jiles impoundments. Such testing can reveal trouble spots such :1S Inyers 0 r lower atrength or excess pore pressure. It cnn tllere[ore m;sJfit considerably in the evaluation of tailings impoundment stnbillty and the planning of changes in construction procedures if required. References Baligh, H.H., V. Vivatrat and C.C. Ladd (1980) Cone Penetration in SoU Profiling, Journ. of the Geot. Eng. Div., ASCE, Vol. 106, No. GT4, pp. 447-461. Begemann, II. (1965) The Friction Jacket Cone as nn Aid in Determining the SoU Profile, Proc. 6th Int. ConI. on SHt.FE, Hontreal, (1/4), pp. 17-20. MINE TAILINGS CIlARACrERIZATION Blight, G. E. (1968) A Note on Fleld VnneTesting of Silty Sol1s, C.~nadian Geotechnical Journal, Vol. 5, 110. 3, August, pp. 142-11.9. .321 Cn'"IHlllella, R.G. nnd P.K. Robertson (1981) Applied Cone Research, Sol1 Necltnnics Series No. 1.6, Dept. of Ciy. Eng., The University of IIritish Columbia, Vancouver, D.C., 30 pp. Gille!'l,ie, D. and H.G. Cnmpnnella (1981) Consolidation Characteristics from Pore Pressure DIssipation After Piezometer Cone Penetration, Soil. Hecllanics Series No. 47, Dept. of Civ. Eng., The University of flrItish Columbia, Vancouver, D.C., 17 pp • IInrr, H.E. (1977) Hechanics of Particulate tledla - A Probabilistic Approach, HcGraw-Jlill, pp. 51.3. Janbu, N. and K. Scnneset (1973) Field Compressometer - Principles nlld Al'plicntions, Proc. 8th ICSHFE, Vol. 1.1, tJoscow. Jones, G.A. and IJ.J.A. V:1n Zyl (1981) TIle Piezometer Probe - A Useful Tool, 10th ICSHFE, Stockholm. Levadoux, .J.N., nnd ~I.H. Dnligh (1980) Pore Pressure During Cone Penetration in Clays, Report No. HlTSG 80-12, tiLT, 310 pp. "arez, I.., Dachelier, H. and Sechet, D. (1976) . Pression Intersti tieUe Developpee au foncage des penetrometers. Proc. 6th Reg. Conf. Europe SHFE, (1,2), S33-S38, Vienna. Peignaud, H. (1979) Surpressions Interstitielles Developpeefl par Ie f,,"coee oous les sols coherent, Canaoion Geotechnical Jnl., Vol. 16, Nov., pp. 8l~-827. Rulnner, W. (197',) Slimes - Dam Construction in tile Cold tIInes of the Allglo American Group, Journ. of the South Afr. Inst. of Hining (, Netnl1. Fehr., pp. 27 /.-281 •• . . Schllll!rtnmnn, J .11. (1978) Study of Feasibility of Using Wissa-type Piezometer Prohe to Identify Liquefaction Potential of Saturated Fine Sand9, Hatervoys Experiment Station, U.S. Army Corps of Engineers, Vicksburg, Hiss. Technical Report S-78-2, pp. 65. l'orstensson, n.A. (197S) Pore Pressure Sounding Instrument, Discus sion, Session I, Proc. ASCE Specialty Conference on In Situ t1easure ",ent of Soil Properties, Raleigll, N.C. Vol. ·2, Pl" I,8-S4. Wisfia, A.E.Z., R.T. ~'artin and J.E. Gadanger (1975) The Piezometer Probe, Proc. ASCE Specialty Conf. on In Situ tleasurement of SoU Properties, Rnleigh, N.C., Vo1. 1, pp. 536-545. APPENDIX Estimation of Penetration Rate for the tleosurement of Drained Conditions A method to estimate an allowable vane silear rate to measure fully drnined shear st rengths of sil ts; was presented by Blight (1968). The sOllie method will be used Here to estimate a penetrntion rate for the piezometer cone for the measurement of · drained conditions. .122 CONE PENETRATION Despite the obvious limitations of the model used here 1.n the cnse of cone penetration testing (such as neglecting expansion of materinl arotmd tip during penetration, etc.), it is interesting to note the conclusions of the analysis. Ti,e .following assumptions are necessary to cnrry out the nnnlysl,,! (i) As the penetrometer is advanced, excess pore pressure 1.s set up within a cylinder of influence of radiu" n. See figure A.1. This cylinder is effectively mnde up of a series of spheres of radius a. n,e pore pressure Js assumed to be uniform within these 'spheres of influence.' (11) The pore pressure on the surface of the 'spheres of JIlf1l1cnce' remains equal to the hydrostatic pore pressure at all times. The surface is thus assumed to represent a druinage surface. (iii) The pore pressure in the individual spheres of rndlus a is set up during penetration of the distance 23, i.e. [rom the time that the cone tip 'enters' the sphere until it 'leaves' the sphere. The distance 2a should therefore be divided by the time for the required drninage to take place in order to obtain the allowable penetration rate.* for the boundary conditions: when t - 0, U - 0 nnd when t ~ 0, u - 0, at r - a (A.I) t cnn be shown that the following solution is v,tlid for pore pre"!ltlre in the sphere wIth radius a (Blight, 1968)! U-l-.! T 1 it (_l)n '''''_ nil 2 2 J [ _ + ~ I: ---- sin - exp (-11 n T) ---8 11 3. n~ 1 n 3· 2 where the time factor, Tee t f v -2- a and U - degree of consolidation in sphere t f - time for required degree of consolidation (A.2) (11.3) One more sssumption is necessary to solve the prohlem; na,uel.y, the relationship between a and the diameter of the prohe. Figure 11.2 presents a plot of T vs. U. "'The assumption of .pore pressure set up over 3 penetration distance of 2a is made here to allow the analysis to be conducted. Reference to figure lIn shows that the assumption of a sphere may not be thnt ullrensonable. MINE TAILINGS CIIARACTERIZATION VOWMEOF INFWEIICE _PENETROMETER .......... SPHETIE OF INfWElICE \ot0.7!101 FI.r.ure 11.1. Volllme of IlIfl .. ellce [)uring Penetration 100 90 eo TO 60 ~ 0 ::> 50 40 30 20 10 0 .1 Cy I, 1=-;;2 1.0 FlJ:llro.' A.7.. Tvf'. IJ ror C"",,,>1I,I;tr' f n n of a !;phcn" 323 ~D 324 CONE PENETRATION Examplli! Find the allowable penetration rate for gold tailings based on the assumptions above and for a" 0.75D, a - D, where I>-diameter of piezometer cone. Solution From Figure A.2, for U - 90?, T - 1.30. Furthermore, from 2 Illight (1968), for gold tailings, average laboratory d v - 600 mm Imino For the cone D ,. 35.7 mm. From eq. A.l for a - 0.75D, t - 1.6 min., which translates into an allowable penetration rate of atout 35 mm/ min. For a ,. D, a penetration rate of about 26 mm/min. is obtained. These penetration rates are much lower than the standard for the testing; namely, 1,200 mm/min. (viii) PIEZOMETER PENETRATlON TESTING CUPT Proceedings of the Second European Symposium on Penetration Testing / Amsterdam /24-27 May 1982 Piezometer penetration testing CUPT G.A.JONES & E.RUST Steffen Robertso.n & Kirsten. Pretoria. South Africa INTRODUCTION Pore pressure or piezometer penetrati:o~ testing (CUPT) has been in use in South Africa on a fairly routine basis for about three years. The -equipment and systems have undergone considerable changes in that period in a planned development programme intended to introduce gradually, more reliability and sophistication particularly in the data logging and processing. The methods in use are briefly described and it is interesting to note the similarity between independently developed systems mentioned by a number of authors in Cone Penetration Testing and Experience, ASCE 1981 and elsewhere. It is hoped that these similarities will allow a consensus to be readily achieved on rationa lisation and standardisation of piezometer penetration testing. The main objectives of penetration testing are to estimate strength, compressibility and consolidation parameters and to permit an adequate description of the subsoil to be made. Some results from three sites recently investigated are given with an indication of the derived design information. Results from these, and many other sites, are combined to enable a soil classification chart to be drawn up based on a relationship between the measured generated pore pres sures and the cone pressures. 2 PIEZOMETER CONE AND SYSTEM Fig. I Piezometer Cone The piezometer cone and friction sleeve currently in use are shown in Figure I. They comply with the European Sub Committee (1977) recommendations. The 600 cone is - 35,7 rom diameter and is followed by a 150 cm2 friction sleeve. A 5 rom wide porous plastic filter is interspersed between the cone and sleeve. A further identical sleeve serves as a housing for a load cell, which measures the cone plus sleeve load; a very similar load cell within the friction sleeve measures the cone load independently. It is of course appreciated that this sytem necessitates a decrease in sensitivity for friction sleeve measurements, as compared with an independent sleeve load measuring system; however the resolution obtainable in strain gauge load cells makes this an academic rather than practical consideration, and it is believed that the simplification in manufacturing and assembly is more than justified. Both loads cells have identical dimensions and have 8 foil strain gauge full bridges. Different load capacity probes are obtained by using load cells with iden tical external dimensions, but with diffe rent axial cable hole sizes. The cone load has a recess into which is cemented a Kyowa PS 10 KA (1000 kPa) pressure trans ducer. The transducer is miniature, measu ring 5 rom diameter by 0,75 rom thick, and the chamber is the minimum practical size into which the transducer ~an be fitted. It may be noted that not one of these transducers has given any problem over thousands of metres of probing. The outputs from the transducer, and the 607 two load cells, are led to the third cylin drical section of the probe which contains voltage regulators, signal conditioners and amplifiers. It has been found that with short cable lengths, this down the hole amplifier system is unnecessary, but under more difficult conditions it has considera ble advantages. The system is arranged so that in an emergency the amplifier can be easily removed or by-passed. The ' overall length of the probe is arranged to be 500 mm, to simplify depth recording . The depth penetrated is measured using a rotating disc, optical linear en coder, driven by the rod movement. The en coder'pulses at about 10Hz , at the standard 20 mm/sec penetration rate, and this con trols both chart recorders and a Sharp MZ80B Microcomputer. The latter has been modified to include 5 A/D converters (3 in use for probing) and gives a continuous display of cone pressure, fricti,on ratio and pore pressure against depth. This data is recorded on tape and later processed on the same system using diskettes and a printer. The probe is calibrated in the laboratory using a conventional triaxial loading frame and cell. The latter has been modified using a special top plate with a 36 mm dia hole with O-ring seals. The loads and pore pressures can be readily calibrated and the effect of cell pressures on the cone load measured. This effect has been discussed by Baligh et al (1981), and is larger than may be generally appreciated when high pore pressures are generated; this problem could be considered as similar to that of in cluding the mass of inner probing rods in the mechanical probing method, i.e. the effec,t is measurable, but only of real sig nificance in particular circumstances. It could of course be accommodated within the software for the data processing system. Several authors, Campanella and Robertson (1981), Tumay et al (1981), Jones and van Zyl (1981), have noted that during dissipa tion tests, after ceasing penetration, pore pressures sometimes increase before dissipating. Although incomplete saturation, or' lag in the measuring system could cause this, the dynamic effects referred to by Schnertmann (1974), at ESOPT I, are believed by the authors to be a more rational expla nation, since the effect has been observed even after the most stringent de-airing procedures. In the calibration system described above, it is very simple to check the relative time response of the cone and pore pressures by introducing a pressure pulse into the water filled triaxial cell. Negligible response time differences were observed, with or without the cone and filter in place, or even when no attempt was made to de-air the filter element. While certainly not advoca ting carelessness regarding de-airing pro cedures, the authors feel bound to remark that up to now they are not aware of having been confronted with problems attributable to lack of saturation ev~D. in tailings dams where the upper few metres are frequently unsaturated. It has however, as noted pre viously, been observed that in a multilayered soil system there is a measurable lag in the field between say a peak cone reading and an equivalent low pore pressure reading. This lag is equivalent to about 40 mm of penetration, and at this stage is believed to be due to the particular geometry of the probe, where the filter element is approxi mately 40 mm behind the tip. 3 FIELD RESULTS Typical CUPT results from three sites are given below to illustrate the different uses of the technique. The sites are a waste ash dam, a mine tailings dam and a river cross ing. 3.1 Waste Ash Dam It was intended that an existing waste ash dam should be substantially raised in height, and an investigation was carried out to measure the strength parameter and the pore pressure regime, so that the dam wall stability could be analysed. The dam measures about 350 m from the ash/water discharge point to the wall, which is about 25 m high. The wall itself is constructed of coarse waste, and is fairly permeaple, showing considerable seepage in the downstream face. Standpipe piezometers had been installed at an earlier stage of construction to monitor the water regime, but these had shown apparently anomolous readings. A series of six CUPT's were carried out in a line from the discharge pipe to the standing water pool at the wall. Dissipation at stops for rod additions was very rapid, so that the ambient pore pressures were easily etablished. These are shown in Figure 2, a diagrammatic cross-section through the dam. A typical cone pressure (qc) result (No 3) is also shown on this section. Table 1 gives the mean qc values, together with the result of sieve analyses on samples taken at ground level at the six CUPT positions - only 7. passing 0,075 mm are shown. 608 I- CUPT ~'----------------------------1.~1 1--------------------------N!-3---380 m N!4 Nt~ Nt6 Nt/ N!2 Uo Uo Uo qc MFa 0 10 20 _ 0 10 20 ~o Uo /0 20 30 /0 20 30 Uo kPa 20 30 0 /0 E i: ! 30 Fig. 2 Section showing pore pressure and typical cone pressure Table I Summarized cone pressures and gradings COPT No 2 3 4 5 Cone Pressure MFa 10 7 6 4 1,5 7. passing 0,075 mm 60 55 88 21 98 6 It would be expected that the coarsest material would be deposited close to the discharge point, and the finest nearest the wall. The above gradings generally reflect this, but the anomaly at No 4 is due to a surface local change due to the construc tion of an access track of coarser material. The cone pressures show a very definite trend across the dam, of higher values close to the discharge point becoming ~ lower towards the wall. Shear box tests were carried out on a number of samples and gave a mean q, = 340 . Various methods of deducing q, and ,5 m : note water table measured at 0,8 m depth, therefore at 6,5 m depth, hydrostatic pressure (uo ) is 56 k.Pa. Measured Ut is 63 k.Pa and therefore Ue is 7 k.Pa. Average cone pressure (qc) in this zone is 4 MPa, overburden pressure (aw ) is about 60 k.Pa, therefore (Qc - noted that the example illustrating the use of the chart IS taken from a project carried out for the Division of Water Deve lopment, Zimbabwe and that their agreement to use the CUPT is greatly appreciated. 9.References BAUGH M.M. and LEVADOUX J.N . (1980) : Pore pressure dissi pation after cone penetration. Research Report N.R.80-U , MIT Dept. Civ. Eng., Cam bridge, Mass . BAUGH M.M., VIVATRAT V. and LADD C.C. (1980) : Cone penetration in soil profiling. J ASCE, Vol. 106, No . GT4, pp.447-46l. CAMPANELLA R.G ., GILLESPIE D. and ROBERTSON P.K. (1982) : Pore pressures during cone penetration testing. Proc.ESOPT II, Amsterdam, Vol. 2, pp. 507-512. CAMPANELLA R.G. and ROBERTSON P.K. (1981) : Applied cone research . Proc . Geotech. Engineering Division. ASCE National Convention, St Louis, pp . 343-362. DE RUITER 1. (1981) : Current penetrometer practice. Proc. Geotech. Eng. Div. ASCE National Convention, St Louis, pp . l-48. JANBU N. and SENESSET K. (1974) : Effective stress interpre tation of in situ static penetration tests. Proc. ESOPT, Stockholm, Vol. 2.2. JONES GA. and RUST E. (1982) : Piezometer penetration test ing. Proc . ESOPT II, Amsterdam, Vol. 2, pp. 607-613. JONES GA. and VAN ZYL DJA. (1981) : The piezometer probe A useful tool. Proc. 10th ICSMFE, Stockholm, 7/19, pp . 489 -496 . JONES GA., VAN ZYL DJ.A. and RUST E. (1981) : Mine tail ings characterisation by piezometer probe. Proc. Geotech. Eng. Div . ASCE National Convention, St Louis, pp. 303- 324. SUGAWARA N. and CHIKARAISHI M. (1982) : An estimation of ' f/>for normally consolidated mine tailings by using the pore pressure cone penetrometer. Proc. ESOPT II, Amster dam, Vol. 2, pp. 883-888. SCHAAP L.MJ . and ZUIDBERG H.M. (1982) : Mechanical and electrical aspects of the electric cone penetrometer tip. Proc. ESOPT II, Amsterdam, Vol. 2, pp. 841-85l. TORSTENSSON BA. (1975) : Pore pressure sounding instrument. Proc. ASCE Speciality Conference. In situ measurement of soil properties, Raleigh, Vol. 2, pp.48-54. VESIC A.s . (1972) : Expansion of cavities in infmite soil mass. J . ASCE, Vol. 98, No. SM3, pp. 265-290. (x) SOIT CLAYS, PROBLEM SOILS IN SOUTH AFRICA PROBLEM SOILS IN SOUTH AFRICA - STATE OF THE ART Soft clays Recent infrastructure developments, primarily along the Natal coast for roads, railways and harbours, have highlighted the significant problems of settlement and stability of embankment and structures on soft clays. Soft clays are not a problem specific to South Africa, and the depositional history, engineering geological and geotechnical properties are similar to those for recent alluvial deposits elsewhere. Nevertheless, there are differences. Probably the most notable is that the alluvial deposits tend to be extremely variable, with sand, silt and clay strata being juxtaposed in a complex pattern both laterally and vertically. This leads to difficulties in ensuring adequate investigation of the strata, which in turn leads to problems of realistically predicting times for settlement. This review discusses the identification, analysis, design and construction on soft clays, giving particular attention to those aspects pertinent to local practice . Introduction Unlike the problem soils described· in other papers in this issue. soft clays have no particularly Southern African connotations and have not therefore given rise to any significant localized techniques of investigation, analysis, design or construction. Nevertheless, the presence of relatively thick deposits of soft clays has frequently necessitated expensive solutions for earthworks and for structural foundations. It is therefore believed that a review of current methods of geotechnical engineering on soft clays is justified on the basis that foreknowledge of the potential problems can lead to economic benefits if the planning, design and construction of projects can be modified to suit the problem. The emphasis of this review is therefore on the description of well established techniques of investigation .and interpretation which are, or Gary Jones, Pr Eng, graduated at Cambridge in 1957 and is now a student at the University of Natal.ljaving worked primarily in road con struction and design in Zimbabwe and Uganda, he developed an interest in geotechnical engineering which brought him to South Africa in 1965 to work in Natal. Spells with consultants and the Natal Provincial Administration Roads Department brought an appreciation of the problems of civil engineering on soft clays which led to his move to the National Institute of Transportation and Road Research. He subsequently moved back to consulting and is a director of Steffen Robertson and Kirsten at Pretoria. He is the author of papers on embankments on recent alluvium and on piezometer probing. Peter Davies, Pr Eng, is a senior partner of consulting geotechnical engineers Davies, Lynn and Partners. He graduated from the Imperial College of Science and Technology, University of London, with a BSc (Eng) (Civ Eng) in 1968. Specializing in geotechnical engineering, he was later awarded the degrees of MSc in 1969 and PhD in 1975. After spending six years working with consulting engineers in Durban, he formed his own . pra~tice in 1981. He has been responsible fOf the geotechnical aspects of . a Wide variety of engineering projects throughout South Africa and also In ~ou~h America, and is the author of various international papers on tOPICS Including the stability of rock slopes and the performance of piled foundations in stiff clays and in weathered rocks. THE CIVIL ENGINEER in South Africa - July 1985 • G A Jones (Fellow) and P Davies (Member) should be, in current use in South Africa, rather than on new developments. It is therefore hoped that this review will be of general interest to civil engineers working in areas of soft clays and accepted that it may serve only as a reminder to specialist geotechnical engineers of the limitations of their craft. Location of soft clays In South Africa soft clays occur predominantly in the coastal areas. The most significant deposits are at Durban, Richards Bay and the Natal North and South Coasts. The Cape East Coast also has areas of soft clays, notably in the coastal strip at Knysna. The southern Cape Coast has soft clays at a number of estuaries and Cape Town itself has some very localized deposits. There are occurrences of soft clays in the inland areas, but these tend to be fairly shallow, very recent transported materials at poorly drained areas such as vleis. It is remarkable how many of these, superficially very poor areas, eventually prove to have relatively simple foundation problems, because the depth of soft materials is limited and the materials themselves are frequently sandy silts and silty sands. This does not imply that there are no inland soft clays, but they are certainly uncommon. Geological Origin The soft clays occur primarily along the eastern seaboard. They are the result of a number of depositional environments which have occurred during periodic changes in sea level. In general the thicker. more clayey deposits have been laid down in lagoon environments of the main rivers. whereas the clayey silts tend to have developed in the tributary systems. However, there are many instances where large thicknesses of soft silty clays occur both in the flood plains of the tributaries and in those of the main rivers. Many of the soft clays contain shells. indicating that these clays were deposited under estuarine environments. During the changes in sea level many of the rivers changed their courses and the positions of the different types of sediments are not now necess~rily ~elated to the present courses. An understanding of the geological history of a particular river system should lead to a better appreciation of the engineering problems; it is. for example. only too easy to make simplifying assumptions about the nature and disposition of the materials within an alluvial deposit and then to devise a solution to suit the si~~le . model. !his may either be uneconomically overconservatlve, If the allUVium is assumed to consist entirely of soft clays, or. on t.he .~ther hand, totally ignore what could be the major problem of variability of the sediments and the consequential differential rates and magnitudes of settlements. Bra~d and Brenner1 describe the wei i-known overseas soft clay depOSits such as the Scandinavian clays, the Canadian Champlain or Leda clays, the Chicago and Boston Blue clays. the Indian and more 355 recently the Bangkok clays. The literature tends to emphasize the homogeneity of these individual. but relatively large deposits. The South Africa soft clays. whilst similar in many respects and amenable to the same analyses and engineering solutions. are relatively small in lateral extent and are generally highly variable. which introduces further problems for the investigations. analyses and engineering solutions. Although of limited extent. many of the soft clays are of considerable depth. up to and even exceeding 40 m in' a few places. which is comparable to the most adverse conditions encountered elsewhere in the world. Definition of soft clays There is no standardized definition of soft clay in terms of conventional soil parameters. mineralogy or geological origin. It is however, commonly understood that soft clays give shear strength. compressibility and time related settlement problems. It is generally understood that soft ' clays are fully saturated and normally consolidated. I n South Africa neither of these conditions may hold, particularly the latter, and such overconsolidation as may exist could be of considerable significance to the solution of a problem. In the near surface clays, which form'a crust. partial saturation and overconsolidation occur together and th~ o~erconsolidation is a result of desiccation due to changes in the water table. In below surface clays, overconsolidation may have taken place when the clay was previously at. or close to, the ground surface and above the water table, but due to subsequent deposition. the clay strata may now be below the surface, saturated and overconsolidated. Overconsolidation ratios of two or three are common and higher values have also been noted. Partial saturation does not in itself cause engineering problems. but may lead to laboratory testing difficulties. Soft clays would generally be expected to have undrained shear strengths from about 10 kPa to about 40 kPa. In other words. soft clays range in strength from exuding between the fingers when squeezed. to being easily moulded in the fingers. or alternatively from being impossible to walk on. to being possible to walk on, albeit with some misgivings. The particle size' classification system approved by the International Society for Soil Mechanics and Foundation Engineering defines clays as those materials with an equivalent particle size diameter less than 0.002 mm. The method of testing to determine the equivalent diameter is hydrometer analysis. In engineering terms. however. materials with as little as about 30 per cent of their constituents being clays, effectively behave as. and are therefore classified as. clays. The plasticity of the material forms the basis of eng ineering classification of clays. The plastiCity is represented by the Atterberg Limits, ie Liquid Limit (wU, Plastic Limit (wp) and Plasticity Index (Ip) where Ip = (wL - wp). The Casagrande A line shown in Fig 1 is often used to define the category of the fine grained material. I n common with many soils. it is the in situ state of the material which is important and not only the nature as defined by the soil parameters measured on disturbed samples. The natural moisture content. (wn). is of particular Significance for soft clays and it is convenient to represent the in situ state by its Liquidity Index (lU, where IL = wn - wp Ip It may be noted that if the natural moisture content is higher than the Liquid limit. the Liquidity Index is greater than unity, a situation which may readily exist in soft clays. and may be of considerable significance in the prediction of the behaviour of the clay on loading. It h{ls been frequently observed that the Atterberg Limits may be signifiGantly affected by the method of sample preparation. viz oven dried. air dried. or not dried at all. Whereas the former is satisfactory for typical road or earthworks materials: structural changes can take place in clays on drying. particularly at oven temperatures. which cause the subsequent Atterberg Limit test results to be inappropriate. Results from tests on samples of alluvium at Richards Bay showed that at one site the Liquid Limits averaged 88 per cent for air dried. compared to 65 per cent for non dried. and that the Plasticity Indices were 51 per cent and 33 per cent respectively. Since the natural moisture contents averaged 83 per cent. the Liquidity I ndices averaged about 0.9 for the air dried samples and 1,5 for the non dried samples. These samples had 356 60 a. 50 )( LIJ a 40 z >- :30 t: 0 I- 20 en :.:: ~ 1·0 ~ ~ ....... Z 0'. 0 ~ to) '" o·~ ~ .............. ~ r--- - 0 to) 0·4 o 20 40 60 100 120 PLASTICITY INO£X Fig 6: Vane strength correction factor for soft clays Field Tests , . Unquestionably however, the rece~t emphasis both in South Africa and internationally has been on in situ testing. The well established techniques are Standard Penetration Testing (SPT) in boreholes. Vane Shear Testing either in boreholes or using a shielded rod system. and Cone Penetration Testing (CPT - formerly called Dutch Probing). A less frequently used in situ technique is permeability testing carried out by installing a piezometer either in a borehole. or occasionally by direct driving oi the piezometer into the soil. The recently developed techniques are the Piezometer Probe and the Self Boring Pressuremeter. The Piezometer Probe (CUPT) described by Jones and Van Zyl" is a development, largely carried out in South Africa. of the familiar CPT. The CUPT system measures cone pressures and. simultaneously, the induced pore pressures during penetration. The nature of the soil can be established from the relationship between the cone and pore pressure (Jones and Rust'2). The option may also be exercised of stopping penetration at any depth and measuring the rate of dissipation of the excess generated pore pressure. This is an in situ permeability test and coefficients of consolidation may be estimated from the results". As earlier emphasized. strata delineation is an essential aim of any investigation; since one of the primary objectives of the strata deiinition is to differentiate between zones of different soil behaviour. for which the consolidation characteristics are vital. then it will be appreciated that the piezometer probe is a powerful tool. Sampling of soft clays without causing significant disturbance is extremely difficult. unless thin walled piston sampling is util ized. If it is not. then for normally consolidated soils, errors caused by sample disturbance. which affect the compressibility parameters more than the shear strength parameters, result in over estimation of the compressibility. This may lead to over estimation of settlements and hence to unnecessarily conservative designs. Conventional sampling also restricts sample size. so that triaxial tests are usually confined to 38 mm diameter samples and consolidation tests to 75 mm diameter samples. Although this is not a problem for homogeneous isotropic soft clays. many of the deposits do have significant layering. resulting in considerable anisotropy. affecting both the shear strength and consolidation characteristics. The ideal situation is therefore to test soils in situ where disturbance can be minimized and sample size i'ncreased. The Self Boring Pressure meter (SBP) was developed for this purpose (Wroth'J). The test is perfo.rmed in conjunction with normal borehole work and consists of adv'l.nci11g a tube into the soil to the test position and then inflating a membrane. The pressure require to inflate the membrane is measured at strain intervals. The pressuremeter operates in a similar manner to a soft ground tunnelling machine in that the soil is excavated at the face and the machine is advanced simultaneously, preventing the yield of the soil into the excavation. The design and operation of the equipment are arranged to min imize soil disturbance. Pore pressures can be measured. and drained or undrained conditions tested. Elastic moduli and shear strengths can be derived. Both the CUPT and the SBP are fairly sophisticated instruments in 364 concept. but are well adapted for field use. In order to obtain the maximum benefits from them it is necessary to closely control their operation to decide what is the most appropriate testing in any situation. This has advantages. since it forces the geotechnical engineer Into the field. and mitigates against the tendency for too much emphasis to be placed on complex analyses. rather than on thorough modelling of the real soil conditions . Although the foregoing emphasizes the importance of in situ testing. particularly as recently developed. it · must at the same time be emphasized that interpretation of the results may depend to a large extent on semi empirical correlations specific to local conditions. whereas the ciassicallaboratory tests do not have this disadvantage. Laboratory tests There have been few advances in laboratory testing of soft clays. Laboratory equipment has become more sophisticated in that manual control of triaxial or consolidation testing can now be replaced by automatic computer controls; data recording and processing of results can also be automated, but essentially the tests themselves have remained the same. In triaxial testing greater emphasis is now placed on the test procedure representing the loading conditions which will apply in the field. since it is now recognized that the stress path significantly influences the soil behaviour. A prerequisite is therefore to return the sample to the same state in the laboratory as existed in the ground before sampling . Complete stress path testing is complex and expensive. and can only be justified if minor refinements in design will produce large cost benefits. A further modification of testing is based on the recognition that few soils are homogeneous and isotropic. This may be particularly so for sedimentary depOSits where the shear strength. compressibility and permeability may be markedly different in the vertical and horizontal directions. Testing is thereiore arranged to model the loading conditions which will apply in practice. Consolidation testing has also changed little for a considerable time. The test has always been time-consuming, but essentially simple. Recent modifications have included automated data acquisition and processing systems. A significant advance has been the introduction oi large diameter test equipment (150 mm) and the facility of permitting either horizontal or vertical drainage. and the measurement oi pore pressures. Fig 7 shows typical results of determinations of coefficients of consolidations from in situ permeability tests and small and large diameter laboratory tests for the soit clay described in Fig 2. The results show a marked apparent dependency of cv, on effective stress and also indicate the large range of c" obtained for very similar samples; it therefore also illustrates the dilemma facing the designer in selecting the appropriate coefficient of consolidation. In the particular case illustrated there is a range of approximately an order of magnitude in cv measured for samples obtained by piston sampling and very carefully conducted tests. It is in these circumstances that the strategy of using piezometer probing becomes attractive. since a large number of dissipation tests can be conducted. Clearly, however. the results of these tests can be interpreted only on the basis of calibration against laboratory tests and experience of real embankment settlements and back analysis. It may be noted that there is often considerable difficulty in interpreting consolidation tests at low pressures because the test data do not always conform to the square root or logarithm time models. This may result in under-assessment of primary consolidation times and hence in an unconservative exaggeration of the influence of effective stress on the consolidation characteristics. Consolidation testing provides both compressibility and consolidation data. The assessment of the former. expressed as either coefficients of compressibility. mv. or compression indices. Cc. is generally accepted to be much more reliable than assessment of coefficients of consolidation. Moderate conservatism in estimating compressibility. and hence settlement. does not usually significantly change the evaluation of the potential problem. or the solution. Estimation of compressibility from conventional laboratory consolidation tests is therefore considered to be adequate. provided that some care is taken to minimize sample disturbance and to correct for any which may occur by the usual methods of interpretation of labOratory data (Schmertman" and Estimation of Consolidation of Settiement IS ). THE CIVIL ENGINEER in Soulh Africa - July 1985 1000 200 100 L- a ~O • >- .... N E > 10 0 5 o G INSITU PERMEABILITIES 0 ROWE CELL (RADIAL> X ROWE CELL (VERTICAL) • 5Omm. CONSOLIDOMETER o 50 100 150 200 250 300 3!50 400 (I' kPa Fig 7: Coefficient of consolidation versus effective stress for various test methods In many cases initial assessments of primary consolidation settlements can be made from the results of cone penetration tests (CPT) . These give moderately realistic predictions for sandy alluvium, but the confidence level in clays is very much lower. Such assessments of compressibility should be viewed as qualitative rather than quantitative, unless correlations of CPT results with laboratory results, or with local experience, are available. There is little published data in South Africa on the measurement of secondary compression characteristics of soft clays. There is of course no difficulty in carrying out the tests; they are simply longer term but otherwise standard consolidation tests. The problem is solely the testing time and hence expense and delay in obtaining data so that the convenience of using semi-empirical relationships dependent on more easily obtained data is attractive. This is unfortunate since the development of a local data base is prevented and no progress is possible. Conclusions Significant deposits of soft clays occur in South Africa, predominantly along the eastern coast. The clays are generally normally consolidated, or slightly overconsolidated, slightly to moderately sensitive and may be very soft. Stratification is often complex and extremely variable, resulting in realistic modelling of the strata requiring considerable investigation. Stability and settlement problems for embankments are common and many structures have to be piled to considerable depths to provide adequate foundations. The costs of civil engineering works in areas underlain by the soft clays are therefore high and investment in proper investigation and analyses is more than justified: It would appear that the South African soft clays do not exhibit any peculiar properties when compared to those described in the in'ternational literature. Investigation and analysis therefore follow normal international practice for these materials. It is worthy of note that over the past two decades many embankments in South Africa have been monitored and a wealth of experience has been acquired. Whilst this has shown that many embankments have performed as predicted, there are numerous instances where this has not been the case. Unexpected secondary compression is one explanation for embankments continuing to settle significantly after the predicted end of consolidation. Undoubtedly laboratory testing programmes should take cognizance of this problem and incorporate long-term consolidation testing. Measurements of pore pressures, such as can be carried out using Rowe THE CIVIL ENGINEER in South Africa - July 1985 Cell consolidometers, add extremely useful information. High quality undisturbed sampling is also essential and it is noted that thin wall piston samplers are available for this purpose. The other. and possibly more likely, explanation for non-conformity with expectations, is that the overall consolidation characteristics as defined by the drainage paths are incorrectly assessed, a problem which is difficult to overcome in highly variable strata . There has been a considerable emphasis both internationally and locally on in situ testing. The piezometer probe is an important addition to the range of techniques which are available. It is relatively low cost and provides a large amount of data on the stratification of the ollen multilayered deposits and on many of the soil properties. Much of this information is of a semi-empirical nature and it would be generally unwise to use in situ testing' to the exclusion of conventional sampling and testing. It would appear, however, that the ideal combination consists of a basic programme of in situ testing, in conjunction with carefully selected high quality sampling and laboratory testing. Techniques to overcome the problems of construction over soft c lays include coating of piles to reduce down drag, sand or geofabric drains to increase consolidation rates and preloading and surcharging under structures and embankments. These methods have become more common and are likely to increase in usage. They are not inexpensive and the warrant for using them must always depend on the prediction of performance, with or without their use, and the relative costs. At present the prediction methods, although soundly based in theory, rely to a l2.rge extent on local correlations and experience. In South Africa this is f2. irly fragmented amongst the different practitioners and therefore the cata base is weakened. It is therefore asserted that the most significant advance in the an of engineering on soft clays in South Africa would be the synthesis o f the knowledge and experience which now exists. Acknowledgements This review is based on information and advice freely given by colleagues !)rs Loudon and Webb. Messrs Clark, Schwartz, Wrench and Yeats. Most of the .:::ata has been obtained from schemes carried out for the Department of Transport. the Roads Department of the Natal Provincial Administration and SATS. The su, :>on of these and other organizations has been Invaluable in our understanding C" ;he soft clay problems and in developing appropriate construction techniques. References 1. Brand, E Wand Brenner, R P. Soft clay engineering, Developments In geotechnical engineering, Vol 20. Elsevier. 1981 . 2. Skempton, A W. The planning and design 'of the new Hong Kong aiq::,:on Discussion, Proc ICE, London. 1957, Vol 7, p 305. 3 . Assouz, A S, Krizek, R J and CorotiS, R B. Regression analysis of $011 compressibility. Soils and Foundations, 1976, Vol 16, No 2, pp 19 - 29. 4. Davies, P. Notes on improvement of foundation conditions by the USE of preloading techniques. Symp on geotechnical processes, University of Nc:al. 1977. 5. Davies, P and Lynn, B C. Design considerations for embankment construc:·,on over soft clays in the Durban area. A E G symp on engineering geology ofC! ~ 'e5 in South Africa. Pretoria, 1981, pp 92 - 95. 6. Simons. N E. General report, Session 2. Conf. on settlement of structu -85. Cambridge, 1974. 7. Mesri, G and Godlewski, P M. Time and stress compressit;, Illy interrelationship. J ASCE Geotech Div, 1977, Vol 103, No GT 5, pp 417 - ":'30. 8. Matsuo, M and Kawamura, K. Diagram for construction contro l of embankment on soft ground. Soils and Foundations, 1977, Vol 17, No 3. pp 37 - 52. 9. De Beer, E E and Wallays, M. Forces induced in piles by unsymmetr ,cal surCharges on the soil around the piles. Proc 5th Europ Conf S M F E, 1972. VOl 1, pp 325 - 332. 10. Bjerrum. L. Problems of soil mechanics and construction of soft clays, Proc 3th Int Conf S M F E, Moscow, 1973, Vol 3. pp 111 - 159. 11 . Jones, G AandVan Zyl. OJ A. The piezometer probe- a useful tool.Proc r Jrh Int. Conf S M F E, Stockholm, 1981 , 7/ 19. pp 489 - 496. 12. Jones, G A and Rust, E. Piezometer penetration testing, CUPT. Proc :nd Europ Symp Penetration Testing, ESOPT II , Amsterdam. 1982, Va' 2. pp 607 - 613. 13. Wroth, C P. In situ measurement of soil properties with the pressureme-:er Ground Profile, 1981 , No 25. ' 14. Schmertmann, J H. Estimating the true consolidation of clay from laboratory test results. Proc ASCE, 1953, Vol 79. Separate No 311. 15. National Research Council. Estimation of consolidation settlement Transportation Research Board. Special Report 163, Washington, 0 C, 1 ~76 . ~65 (xi) PIEZOCONE TESTING TO PREDlCT SOFT SOIL SETTLEMENT G60technics in the African Environment, Blight et al. (eds) e 1991 Balkema, Rotterdam. ISBN 90 54100079 Piezocone testing to predict soft soil settlement G.A.Jones Steffen, Robertson & Kirsten Inc. , Pretoria, RSA E.Rust University of Pretoria, RSA ABSTRACT: Piezocone testing is a versatile and comprehensive geotechnical investigation technique which is particularly suitable for recent alluvial deposits. In common with most in situ methods it is essentially a semi empiricial system for which the results require to be correlated with conventional geotechnical parameters measured by other means. Such correlations are described here for three major road embankments on the Natal coast which were constructed in the late 1970's and have both extensive initial laboratory test results and subsequent monitoring data. The correlations allow constrained modulus coefficients, CEm: and a cone time factor to be obtained from piezocone data so that coefficients of compressibility, I'I1vo and coettlcients of consolidation, Cy. can be assessed for these alluvial deposits. The resultant CErn' 1,24, for these subsoils is shown to be significantly different from that quoted in the literature. This may be because the subsoils were very highly stressed and further research on this aspect is necessary. 1 INTRODUCTION Over the past two decades the development of the national road system has led to a large number of high embankments being constructed over soft deposits. This is particularly so along the Natal coast where there are approximately forty freeway crossings of rivers, of which thirty involve a significant extent of alluvium. For reasons of road geometric alignment and provision of adequate flood capacity, the river crossings generally result in embankment heights in excess of 7m which often gives rise to settlement and stability problems. These problems are never unforeseen and considerable efforts have been made to predict the behaviour of embankments, since this has a major impact on the satisfactory performance of the road as a whole. In practically all cases the stability has been adequately ensured using the conventional geotechnical approach of undisturbed sampling. laboratory testing and limit state equilibrium analyses. Where stability has been shown to be a potential problem it has been overcome by anyone of, or a combination of the techniques of flattening side slopes, adding berms, installing subsoil drainage and controlling the rate of construction'. Compared with stability assessment, success with the prediction of the amount and rate of settlement has been less impressive and, it may be added, this is not a problem confined to Southern Africa. In order 283 to address the problem the authors, amongst others, have developed over the last 15 years the technique of piezocone testing (CPTU - cone penetration testing (CPT), with pore pressure (U) measurement). This is now a well established procedure internationally, albeit the theoretical modelling and interpretation are not as well developed as the sophisticated measuring, recording and data processing techniques. To a large extent interpretation remains semi empirical in that shear strength and consolidation parameters are obtained by comparison of CP11J results with other testing techniques or with performance measurements. Three road embankments along the Natal coast, viz Umgababa, Urnzimbazi and Umhlangane, Figure I, have been extensively monitored since their construction in the late 1970's. This has provided a unique opportunity to back-analyse their performances to obtain settlement and consolidation parameters. A research project (Rust and Jones, 1990) was sponsored by the Research and Development Advisory Committee (RDAC) of the South African Roads Board, which enabled the back-analysis and piezometer cone testing to be carried out. From the latter, settlement and consolidation parameters were derived and compared with the measured performance data so that appropriate correlation factors for these alluvial materials were obtained. This paper summarizes the fmdings of the research project and discusses the implications. Fig 1 Site Plan 1.1 Geology rt!>....;.;~_ UMGA8ABA I I @ The geology of the Natal coast has been extensively described by Brink (1985), King (1942), Orme (1974), Maud (1968), Moon and Dardis (1-988) and at the particular sites of Umhlangane (Sea Cow Lake), Umgababa and Urnzimbazi by Jones, Levoy and McQueen (1975), Jones, Rust and TIuczek (1980) and Jones and Rust (1981). The coastal. development of Natal has been predominantly along a narrow, relatively flat . strip fringed by dunes of Berea red sands. Immed~a~ely inland, hills of shales of the Ecca Group and tillites of the Dwyka group of the Karoo Sequence are encountered. The hinterland topography becomes much steeper and more deeply incised. Further inland sandstone of the Natal Group (locally referred to as Table Mountain Sandstone) of the Cape Supergroup and granite occur. The climatic N-value in the area where the rivers rise is generally 2 or less and rocks are deeply weathered and readily eroded. The N value as defmed by Weinert (1964) is calculated from the evaporation during January, Ej, and the annual precipitation, Pa, viz N 12 Ej/Pa Hence where N is 2 or less the climate is hot and humid and rock weathering is mainly due to chemical decomposition. Due to changes in sea level the valleys have become drowned in the lower reaches and are characterized by lagoons with deep alluvial Tabl. t. Labontory lest results .. inUalOfJ AaI!.A/OEI'O!U r ~IAn!R'AI~ ~1AY ~I.T UOllin nA~IIC: MUlsn lkp. nr;,SCltlrTIO'" <1: 1,_ 2 (11_ IJMl r I.lt.IIT CON"",," "I. (';c, ('-' ('" ('" ('" St-lltc .... f1Dodrlt- :.JcM (fer"""tr fb, II )I )Jl II )l UK ...... ' Illad .... ,ria' 2J " ., 2J " D .... e.t • .,. JoIi4·r:rCT"'ct., 1O ., " )I )I ~'''' OlM..~,cb, OJ )) " )) .. II.klt .... ".,. e. ...... U, ... S'c,.....t ii'lyd., II 1O .. .. '" r ..... n-. D .. ll'c,,,,lt, cit!" 2J II .. lQ )I '-01 ILM" -*, d.y II (0 " )) ., Tlbl. 2. Labora(Q..,. test result1 • she>( strength AIU!A MATERIAL SIIEAR STRENGllt rARAMETERS OESCRlrTlON .. , . U'. \r. s...t~C- 811ck Wty da, 15-25 1l 0_ Dlac.k Jikyd.,. 11).1l 1l aa... S .. dysilryda, IS-l5 Il).lll 1J.lO 8., 8IKkt.iJtydl, Il).lll S·1l 2:l-1l TJ.ble 3. Laboratory (est results .. cousotidJ,uoa. AREA MATEIUAL OESOUPTlON COEFflClE"TS OF COMi'RESSIBILITY AND CONSOUDATION 284 ( .. 1~ ( .. 1;;"') Soolhc:.,.,. Lipr Vey sudy s.ilry da, 0'>-1.0 5-10 BI.d siJrydl Y 0.501.5 I·S Dutbu Blick wry w y O.s-~O 0.5·2.0 R.icb.ards Bay U, btVc.,s.acd w ty d.lY 0,1-1.0 5-10 B j,Jci;Jil~dl :' 0.5·1.0 I·S deposits. The sources of these deposits include granite, sandstone and shales and this results in multi-layered deposits of sands, silts and clays. The larger rivers viz Tugela, Umgen~ Umkomaas are deeply incised, have depths to bedrock of 50m or so and are predominantly sandy with basal boulder layers. The shorter rivers are much shallower, and generally have alluvial deposits of about 20m of sandy silts and clays. The recent alluvial deposits are normally consolidated and laboratory testing does not show unusual characteristics. On the flood plains the surface deposits are often lightly overconsolidated due to desiccation. The soft clays which are the source of the major embankment problems comprise very soft to soft, slightly organic, medium to highly plastic, mid to dark grey, silty clays often containing shell fragments. The liquidity indices are often very high and the clays may be regarded as typical young clays. At two of the sitt!s, Umgababa and Urnzimbazi the subsoils are heterogeneous, having layers of intercalating silty sands and clays. At Umhlangane, however, the subsoil is a practically homogeneous very dark grey silty clay for the full depth of about 18m. Typical laboratory test data from the original site investigations and Brink (1985) is given in Tables 1, 2 and 3. 1.2 Piezocone Testing (CPTIJ) Piezocone testing has been described in many publications and perhaps the most useful recent compendium of information is that by Meigh (1987). The piezocone is distinguished from conventional cone penetration testing (CPT) by having a system for measuring ~he pore pressures generated by penetration of the cone. This data can be interpreted to considerably enhance the identification of the subsoil material type, the shear strength and the consolidation characteristics. A brief review of the relevant interpretation methods is given below. 1.2.1 Subsoil identification A soils identification chart, Figure 2, has been drawn up by Jones and Rust (1982, 1983) based primarily on. data from South African alluvial deposits. In this chart the dynamic pore pressure, ud' viz the total pore pressure uT measured during penetration, minus the hydrostatic pressure, uo' ie uT - uo' is rela ted to the normalized cone pressure ('Ie - avo), viz cone pressure. 'Ie minus the effective overburden pressure. avo. 100 [2] Cl.,. CEJ] Cia,.., Slit EJ Sondy Silt . OSO"cS. r-~?'t--t"'=':"""--+"':"::'::.:'i.:::.",::-::o.~".::,,+-__ . ". Fig 2 Soils identification chart 1.2.2 Undrained shear strength The undrained.shear strength is derived directly from the cone pressure, CJc. by using the following equation: 285 'Ie CuNk + avo where cone pressure undrained shear strength cone factor effective overburden pressure. 1.2.3· Effective shear strength Various methods of assessing effective shear strength parameters from CPTIJ data have been presented by a number of authors and are summarised by Meigh (1987). The application of these techniques are fairly complex and fall outside the scope of this paper. 1.2.4 Compressibility The conventional coefficient of compressibility, Illy. may be assessed directly from cone resistance by the following equation: lIly 1/ IXm'le where IXm 'Ie constrained modulus coefficient cone resistance Values of IX for a variety of soil types are available in the publisY{ed literature and summarized by Meigh (1987). 1.2.5 Consolidation Coefficients of consolidation, Cy. can also be obtained from CPTIJ results by relating them to the time taken for excess pore pressures to dissipate after ceasing penetration. Although theoretical solutions have been suggested for this relationship, Jones and van Zyl (1981) put forward a semi-empirical re~ations~ip for the Natal recent alluvial deposits. In t~lS the tune for 50% piezocone dissipation, t50, was dlrectly related to Cy on the basis of comparing these t50 with laboratory consolidation test data for . samples taken from the same positions as the piezocone dissipation tests. Thus a cone dissipation factor was determined with a numerical value of 50 in the following equation: Cy 50/t50 where Cy coefficient of consolidation, m2/year. half piezocone dissipation time. minutes. 2 PROJECT METHOD AND INVESTIGATION 2.1 Method As stated previously comprehensive investigation and monitoring data was available for a period of longer than 10 years for three major embankments. Three information sets were available viz the original site investigation data, the subsequent monitoring information and the recent piezocone tests. Using the monitoring data from the Umgababa embankment, the field parameters derived from the back-an~lysis were compared with the new CPTU information to obtain the appropriate correlation factors, ie cxm and the cone dissipation factor. These factors derived from Umgababa were then used with the CPTU data to predict the performances of the Umhlangane and Umzimbazi embankments. The CPTU predicted performances were then compared both to the measured performance and to the original investigation data in order to check the efficacy of the methods and factors. 2.2 Investigation A total of 21 CPTU's were carried out in June 1989: 11 at Umgababa, 6 at Umzimbazi and 4 at Umhlangane. Of these, 6 were put down through the medians of the embankments. These therefore included the effect of the load on the subsoil. The remainder were either close to the embankment toes and hence also included the embankment influence, or deliberately away from the toes to exclude this effect. The CPTU's were carried out using recommended international procedures and equipment. Dissipation tests were carried out at frequent intervals. Overnight dissipation tests were included to obtain the full dissipation records and also to defmitively measure the ambient pore pressures under the embankments. These pressures were known to be in excess of hydrostatic since consolidation was not yet complete even after more than 10 years. Using the soils identification chart and the borehole logs of the original investigations, geological sections were drawn up so that strata thicknesses could be accurately defmed for settlement analyses. 3 ANALYSIS AND DISCUSSION 3.1 Umgababa Parameters The subsoil compressibility and consolidation parameters for the Umgababa embankment wero:: evaluated from the monitoring records. Information on the ambient pore pressures was obtained from the CPTU. Figure 3 shows the ambient pore pressure against depth and very clearly indicates that the clay layer had excess pressures. 286 2~~--__ ----------~-------' a a.. ~ zoo g I~O I&J ! en en I&J 100 a:: a.. I&J a:: o a.. ~O a.AY LAYER \1/", I f\ I , , 20 , \ , \ \ \ , , , '~ 10 DEPTH (ml -x- AMBIENT PORE PRESSURE HYDROSTATIC PORI! ~RUSURI! o Fig 3 Umgababa : pore pressure v. depth The detailed geological sections were obtained from the CPTU's and original investigation data. Figures 4 and 5 show a typical CPTU log and section. From this data and from the construction and monitoring history it is possible to represent the embankment performance using a classical two dimensional consolidation model. The steps for this process are summarized as follows. At the cross section, Figure 5, the total thickness of clay is 14,2 m and the drainage path length of the major clay layer is 6,55 m. The embankment design height was 5,6 m, but because of ongoing settlement during construction (1,1 m) the fmal thickness of fill was 6,7 m, hence the increase in total stress was 147 kPa (6,7 x 22). The large embankment width to subsoil depth ratio results in there being practically no decrease of stress with depth. The ambient pore pressure at IS,S m depth was measured as 205,8 kPa and the water table was at 3,3m. Therefore the excess pore pressure, u , at this depth of 7,0 m into the clay, ie just below 1he half depth, was ue 205,8 - (15,5 - 3,3) 10 83,8 kPa Hence the degree of consolidation, Uz' at that depth was: Uz 147 - 83,8/147 0,43 Using the Tenaghi consolidation theory and Taylor's chart relating degree of consolidation, depth factor and time factor TV' the latter has a value of 0,33. From the Tenaghi rel~tionship between Tv and percentage consolidation U, it can be shown that o was 64,2% at the time of the CP11J investigation, at which stage the settlement was 2,69 m. -COIC~A~(:WIJ'.' OT-__ ~~~~l_. ____ ~ __ ~"_. __ ~gw. 10 11 H -- I. .. o 100 zco XK) 400 .. ,. - P'QIII( PlftUSUM I~I Fig 4 Umgababa : typical piezocone log koa . IJ ,ORO Fig 5 Umgababa : geological section 287 The present rate of settlement was also known, 7 to 8 mm/month, and the problem therefore becomes one of trying various combinations of Cy and my so that the best fit for both present settlement and present rate of settlement are matched. A spreadsheet approach was adopted to enable multiple iterations of combinations of Cy and my to be performed and Figure 6 shows a realistic fit of one of these combinations to the settlement record. This gives the following compressibility and consolidation parameters: Cy 1,5 m2/year my 1,8 m2/MN. The fit is not perfect in that the model predicts a 1989 settlement rate of 9,5 mm/month, whereas the actual is 7 to 8 mm/month, and also the measured settlements in the early stages are more rapid than predicted. However, no attempt was made to account for stress dependency of Cy on effective stress, or for changes of drainage path length with time, since it is believed that these complications would not add anything useful to the analysis. These back-analysed Cy and my values are in agreement with the laboratory test values given in Table 3. The model predicts that the 90% consolidation will take a total of 24 years and that the total settlement will then be 3,86 m ie a further 1 m of settlement has yet to occur. SETTLEMENT RATE (mm/month) ~o 20 10 I. +----''------'-----'---1 .. .., c; o 10 !l • o ~ I- en > -;; -f ,.. • 5. • < , .. j -G ' .~ . .. 1 -t j -, _,t :f ~ i 'I ,." '" 1 L ~ ! .' :i , Proceedings of the Eighth Regional Conference for Africa on Soil Mechanics and Foundation Engineering / Harare / 1984 A comparison between in situ and seismic methods of site investigation in the alluvial beds of the Sabi river, south eastern Zimbabwe C.E.REA & G.A.JONES Harare. Zimbabwe SYNOPSIS: Site investigations for subsoil conditions can be done by means of diamond drilling, in situ probing and geophysics. A site investigation in the alluvial beds of the Sabi river required piezometer cone penetration testing and seismic regraction surveying along two section lines to determine the nature and depth of alluvium as well as the profile and consistency of the bedrock. The two methods are summarised and a comparison is made of their advantages and limita tions, using examples from the investiga~on. It is concluded that the two methods complement each other and that a combination of both techniques was necessary for this investigation. INTRODUCTION The exploration of subsoils for a structure can be done directly by diamond drilling, by in situ penetration methods or indirectly by means of geophysical methods. The choice of method or combination of methods depends upon soil conditions, information required and economics. An investigation of the Chitowe Dam site in south eastern Zimbabwe required ~~e combination of piezometer cone penetration tests and seis mic refraction teChniques to evaluate the sub soil and bedrock profile on two sections across the alluvium comprising the river bed. This paper summarises the piezometer cone penetra tion test (CUPT) and the seismic refraction method, illustrating the merits and limitat ions of each in relation to the investigation of the section lines at Chitowe Dam site. PIEZOMETER CONE SYSTEM One of the most widely used in situ testing techniques is the cone penetration test (CPT). In recent years the CPT equipment has been developed to monitor continuously pore press ures and penetration resistance. (Jones and Rust, 1982). The analysis of the CUPT data enables the identification of subsoil in addition to the parameters obtained from the conventional test. The method was developed semi· empirically by correlating CUPT data with independent soils classification tests. The piezometer cone is basically the standard 35,7 mm diameter, 60· cone (Figure 1). ~train gauge load cells measure the cone press ure and the total cone plus sleeve friction enabling conventional friction ratios to be deduced. A filter ring behind the cone enables the pore pressure to be monitored by means of a miniature transducer. The two load _cell and pore pressure transducer outputs are amplified in the penetrometer tip and transmitt ed by cable through the rods to the surface. The surface control, wi~~ variable gains, Fig 1 Piezometer Probe 39 SLEEVE lOAD CEll . PORE PRESSURE TRANSOUCER POROUS FilTER U. IkPo 400 zoo 100 Fig 2 CUPT Test in Sabi river d,ns. '\ .... ,'1 din .. ' i~)lO .9.C..5.L --- _~ I MPaI Fig 3 Soils identification chart conditions the analogue signals and feeds them into a chart recorder and a data acqui sition unit. An optical shaft encoder senses t he depth of penetration (Figure 2). A chart has been developed correlating the excess pore pressure (U ), and the cone press· ure (Q ) minus the total vertical stress (0 i c ~' against the soil type and consistency obtained from laboratory tests of undisturbed samples and examples cited in the literature (Figure 3). The interpretation of the results of a CUPT test by means of the soils identification chart is shown on Figure 4 . A detailed soil profile can be obtained in this manner to considerable depth in a matter of several hours for each test, The depth to bedrock is obtained directly by refusal of the probe . SEISMIC REFRACTION SURVEYING This method of geophysical exploration is well documented in the literature (Hawkins, 1961; Mooney 1977; Redpath, 1973) and has been used frequently in shallow geo technical investigations. The refraction method consists of measuring the travel times of compressional waves generated by an 40 impulsive energy source. The energy is detected by means of geophones and recorded on a seismograph. The instant of the shot is alsc recorded on the raw data which consists of travel time and distances. The time-distance information is then manipulated to convert it into the format of velocity variations with de?th. In field conditions where there are discontinuities and variations in the soil profile it is necessary to use the continuous seismic refraction method to determine reversec seismic profiles. The traverses are arranged such that overlapping first arrivals from opposite directions are recorded for each refrac~or at each geophone location. This usually necessitates a minimum number of seven shots per array of twelve geophones, but depends upon the depth to the hardest refractor , the number of intermediate refractors, and the spacing of the geophones. The total length of each traverse is selected to suit site conditions and is normally a multiple of about four times the depth to the hardest refractor. Various methods of interpretation are used depending on the amount of data available and the particular requirements of the o o o 5 15 100 10 200 20 300 Ut (kPe) 30 qc (MPe) ._- PORE PRESSURE{UI) CONE PRESSURElqc) '-'-" 4..'} .~.-" J..,"'-. _._-_._._. -_. __ .--...... ___ .J. LOOSE SAND FIRIoI SILTY CLAY SAND LENS FIRIoI SILTY CLAY SAND LENS FIRM SILTY CLAY CLAYEY SILT IoIEDIUIoI DENSE SILT FINE SAND .- .--.- SOFT --------~ · ----·--ROCK Fig 4 CUPT test interpretation investigation. The commonly used "delay time" or "time depth" methods are described in the literature and prove to be relatively accurate if there are few abrupt changes in bedrock topography. The "delay time" is analagous to the time taken by a pulse to travel up or down ward through a layer from one interface to another. If the delay time for each layer below a particular detector can be determined, then the depth beneath the detector to eacL layer can be computed. Thus by means of a continuous reversed time-distance plot over the full length of the traverse, depth information can be computed at each detector. The true velocity of the longitudinal waves for each layer can also be obtained from the continuous reversed time-distance plot. This is related to the consistency of the material and with prior knowledge of the geology, an assessment of rock quality can be obtained. .The range of velocities for various materials has been documented in the literature (Jakosky 1957). 41 COMPARISON OF THE IN SITU CUPT TESTS AND THE INDIRECT SEISMIC REFRACTION SURVEY Two section lines across the alluvial deposits forming the bed of the Sabi river were investi gated by both methods. The CUPT tests were carried out at 25 metre intervals and geophone locations were at 5 metre intervals on sand banks, and 10 metre intervals over stretches of flowing water. Figure 5 shows the ~ections with the results of the seismic refraction survey superimposed on the detailed results obtained from the CUPT tests. The sections show the nature of the bedrock profile below the alluvium that now forms the river bed. A deep paleochannel is evident on both sections and is associated with a doleitic intrusion into the sandstone country rock as confirmed by diamond drilling on a section line. The longitudinal wave velocity of the bedrock in the paleochannel zone was found to be low, which indicates a higher degree of weathering or fracturing. The CUPT tests show that the alluvium consists predominantly of free-draining loose to medium dense sands with ·some layers and lenses of soft clays and stiff silt&. . The detail from each CUPT test is correlated to construct an idealised section of the soil profile which has value in assessing the general nature and properties of the alluvium with a good degree of confidence. It is evident that on Line A the loose sands extend ed only to about 10 metres depth across the full width underlain by soft to firm silty clays, stiff to hard sandy silt and in the central channel, medium dense silty sand and firm to stiff silty clay. Line B, however, is shown to have clays and silts near to the banks only, and one isolated silty lens in the central channel, the remainder being sands. The seismic refraction results are more generalised and identify only two refractors which are the saturated alluvium and the bedrock. The alluvium is only evident as a homogenous layer of constant velocity which has a slight variation across the sections. No identification of intermedi~te refractors is possible due to the limited lateral extent of the thin silty layers and lenses except on the left bank side of Line A. On this portion of the section it is interesting to note that the main refractor was identified at a relatively shallow depth, having a fairly low longitudinal velocity. When correlation with the CUPT tests was made it became evident that the refractor was coincident with a continuous layer of stiff to hard clayey and sandy silt about 5 metres above the true bedrock level as shown by refusal of the CUPT tests. The true velocity of the hard sandy silt was computed as 2950 m/sec which falls within the range of velocities for sandstone material. The under lying materials were shown by the CUPT tests to be medium dense sand and soft clay. This condition is termed "velocity inversion" and -1>0 N • • • • 4 • • _ ... . _, · • • , ';" • •••• t~. ~ .',.~ • • , ~t.ti " .. j' ' ..... ~ ... . ... . " C514 1 .... ~'Dub, J/7 ---~" ......... .' .. 1400 400 ~ .. ----........... m U5 I C52 C5 4 C5 CH C51 C51 C5 350 C510 C51 CSI1 CS .................. - - .. - -- - .J,... -- - ----- - - - - - - --.1 I I I ltO S H lOOSE SAND 1.1 ...... f'NE SllTYC !..------z1:!·~ _~NDY SlL T LOOSE A~I\F SANOY SILT LOOSE SAND . "'=C ~ - ~:::CT. .. - -";;11Olr-Ir.~1I:1l.~:':JE=-=---- -_ -::." ____ _ _ _ __ liS ~_ ~::5it..:;~ ---r- - ~~~t: ~~ .. .:; _---vrAvS IFF ........ - -~-STIF To~;;~' -__ • __ ~._ .... - -.~~:.:t:?::[ -:.-- V~;;ST' f-~".~O t;',;ys'~r-- CLAYE SIlT C'lAVE 5'~~ 110 "T OCl"tf;Yi ----- r~Joru ... _ C ENSE -__ . "70 _~ _-- I HARD CLAYEY SIlT..J ~T 50fT CLAY - ..... SAND '1EO'U'" DE ~E .. ro(DIIDOOE SAIII! STIff SANOY SILT 375 )7S .." _ ..... ~ILTY SAN ti~ Y. HARD SANDY SILT I 1400 . ~~ " ___ - .... ~~OBE REfUSAL LEVEL no 11 20 , .. ~ -'~l4-O£N r- rr '~~' ...}ANO __ I; 3600 8EORO~ .~ PRI'Fllf-----\ • WsV'PRii1lii -~ _ us r SE I~MIC , \, __ • SAND " 20'0- - 1"0 LEVEl , VERY STI~I .... ~ SOFT SANOY MEDIUM DENSE ,Lt --- 4440-- ~----..- .. ---.--' ......... ~NDY ~tr,oo~~A:AHD SAND MEDIU OfN SAND ~~"ilJ ~.1S00-~ - _ .. I jH PROBE REFUSAL LEYEL 3410 __ ~E_~~~ iSf So NO ;"" "F"" --- U,o--- ~ - .. ~O~ SAND SAND CLAY_ ft -4440- \ -.= MEDIUM DE~ BEDROCK PROFILE I -~ AND m ,J;Ro ~ (Sf'S""C) , ., , .~ ~-1. 1'---~4S0 _ _ _ -~000------1 \ 1h I --' 11ts )to t----+- , ' 1 _ ___ L__ ..L...:. al_"l..,_al~j~dP'"' · ~ I ~-- , JtO liS , .It > , I· .. · ... 1 ____ 1 L-V ' .. ~ ~ _ ~'1I 1 ""'= I liS l80 , d~'.p-"' ... "''' "'_.,<1 .. ~ .... o;."..., ~""U £....1' 'J.O 17S , ..... -~ ~.... I r= -- .. mnx ;>Any Ml:lIlUM ytN)1: '" 1]75 )10 I \ ' :f). nt"),l~_J_"'''~'''''''' ..... '~-t"'J~ '170 US I ..,., ~__ 'lU )60 I ~ .J' 'no .......... _.1' ........ _-- lSS , 15S LINE II SECTION ACROSS RIYER BEO 230m . NORTH OF CENTRE LINE Figure 5 Sections across Sabi river comparing resul t s from CUPT t e s t s and se ismic re f r action survey. .::·t 41 ~: :::: ~ '" • ~ 'ii L j ~ J J . ~ '. ~ the refraction method is unable to detect the presence of the lower velocity layers. he second limitation is that the refraction T thod cannot detect thin layers or differen me 1 ' f th . oor tiate a refractor accurate Y ~ ere ~s ~ velocity contrast. The many thin lens~s.~ ection line A amply illustrate this l~~tat­~on whereby the refraction method indicates a homogenouS medium. The lack of velocity con trast between the alluvium and the sandstone bedrock is clearly evident in the paleochannel on both sections and hence the inaccuracy of the seismic refraction profile in these port ions. The velocity contrast should be a factor of two for reasonably accurate depth predictic tions or even to enable detection of an under laying refractor. This offers an explanation for the inability of the refraction method to detect the underlying sand~tone on Line A below the hard sandy silt layer of velocity 2950 m/sec because the velocity of the sand stone would have to be greater than about 6000 m/sec which is far above the normally accepted limits cited by the literature (Jakosky, 1957). This limitation is referred to as the "blind zone", and is only eliminated in conditions where layers have thicknesses of the same order as the geophone spacing and where refra ctors have a velocity that contrasts sharply with the overlying material. A discrepancy in the seismic results indicated a step in bedrock topography between CUPT tes~ CS5 and 6 which correlates with the point at which the hard sandy silt layer merges with a zone of medium dense sand, thereby accounting for the sudden loss of refractor continuity. A similar discrepancy was also evident on section Line B between CUPT tests CN3 and 4 and is explained by the presence of a very dense to dense sand layer of limited extent. These examples serve to illustrate certain limitations in the seismic refraction method. The seismic refraction sections illustrate that the true velocity of the sandstone varies considerably. The true velocity is directly related to the consistency of the material and hence the rock quality can be estimated throughout the section. This information is of value to the engineer and geologist in assessing the probable areas of incompetent bedrock. The low velocity zones in the refractor are usually associated with major jOints, faults, fracture zones or intrusions in the bedrock which can be explored more intensively by means of diamond drilling. It is obvious that the CUPT test indicates nothing about bedrock quality and depth information is only accurate in boulder free deposits. The linear continuity of bedrock tOpography is assumed between each CUPT test which implies that close spacing of tests is a necessary requirement for obtaining accurate profiles. CONCLUSIONS The piezometer cone penetration test is an in- .,. "'~ ,". " . . situ test for soils which can be used to deter mine the consistency and type of soil by correlating cone pressures and pore water pressures using a materials identification chart. The method gives detailed soil profiles at each test position which can be used to obtain idealised geological sections. The method is limited to saturated field conditions which somewhat curtails its use in drier regions where partially saturated conditions are more common. The test is limited also in the inabi lity to differentiate refusal due to obstruct ions such as boulders from bedrock level. The tests are conducted quickly in a matter of a few hours to depths of up to 35 metres, but the cost is relatively high due to the specialised equipment and personnel required. The seismic refraction method is an indirect technique of obtaining subsoil conditions in terms of the depth and true velocity of the refractors in the vertical section. The true velocity is correlated to the consistency of the materials to give a generalised section showing the profiles of the refractors and the consistency. The method can be used in most fielq conditions and is not influenced by small boulders or obstructions. The method has the advantage of enabling 'an assessment of the bed rock quality to be made continuously across a section. The method is limited in the inabil ity to detect low velocity layers underlying higher velocity layers and in the requirement for good velocity contrast between layers. The surveys can be conducted quickly covering up to ten traverses per day, making it cost effective in most site investigations. The two methods of investigation complement each other in several ways. The depth of bed rock is measured accurately at piezometer cone penetration test positions but the seismic refraction method provides more continuous information with accuracies of about 10% except where it is limited by velocity inversion and poor velocity contrast. The penetration method gives detailed information on subsoils to bed rock level and the refraction method gives fairly detailed information 'on .bedrock quality and co~tinuity. Thus it is illustrated that "both site . . investigation methods provide sub- 43 soil information; but that' the combination of the two methods is more effective in present ing a more complete understanding of the sub soil conditions. It must be stated, however, that both methods are indirect and that they should be confirmed by diamond drilling. ACKNOWLEDGEMENTS The data has been extracted from a comprehen sive report on the geotechnical investigation for Chitowe Dam site for the Ministry of Water Resources and Development, Zimbabwe. Particu lar thanks are expressed to Mr T C Kabell of the DeSigns Branch (M W R D) for encouraging the use of seiSmic exploration techniques and to S R K and Partners for providing the data. REFERENCES Jones, G.A. and E. Rust (1982), "Piezometer Probe (CUPT) for subsoil identification". Int. Symp. Soil and Rock investigations by in situ testing. Hawkins, L.V. (1961), "The reciprocal method of routine shallow seismic refraction investigations", Geophysics 26, pp 806-819. Mooney, H.M. (1977), "Handbook of engineering geophysics", Bison Instruments, Inc. Redpath, B.B. (1973), "Seismic Refraction exploration for engineering site investiga tions" u.S •• Army Engineer Waterways Experiment Station. S . R.K. and Partners (1982), (No. 79/1) "Report on the geotechnical investigations for Chitowe Dam site, Chisumbanje". Jackosky, J.J., "Exploration Geophysics", Trija, 1957. 44 APPENDIXll CONSOLIDOMElER-CONE RESULTS APPENDIX n CONTENTS Figure 11.1 TSPC Series 1 4 Ib " II.2 TSPC Series 2 SIb 11.3 TSPC Series 2 11 Ib 11.4 LPC Series 2 11 Ib II.S TSSH Series 1 1 Ib II.6 TSSH Series 1 7 Ib " II.7 TSSH Series 1 11 Ib I - I --,( ---( --( - --{( --,( ( --\')-< o«)~~6 ------1( - --( ------1 --,( ---{( -- ('----Ie - --( --( - --I( --( _ ---{ --( .--( - -( --( - ( ---{ - - ( - -( ----( --{ - ---Ie -~( ----Ie ----I( -- --, ~i1 ~ ~, t~:TI. tFF~-~i! ~~_~~E; ==:=:f_=_=-=ti~tt1I-~~~==t-It- ~==~i=- ~=---+-~_F~~=_r===r===~F; ==. =1= I, ~=t+_=._==l=_ =r~=F=~I--~:-=' ==~i--~T==--~=-=--~=f=-=I=- r--=tt===fl =+I=-~-1-==-r~-r==~1-=--->-I!==r; "'.:-=r-==+ =: = .. -=l=1-=~-=-r1 ~=I~~-_=r==- -:~ --/ r J \-- l~,- r +~ _ = ~~r - - 1---1 l ~ I I ! :\ =-+ 1- -1 I I; 4 -- ---t -! --t, ! 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I---::::-_--~ ~~~ ~ ~/1~75 ::::~::::- -=---~ . I~Ol~'::-+-.......::::~~--.-::~:~lt - :--1--r i-,· :--==.- ;- --~ 1---~+ !:=----~ I--~- ,- ~~ ~: = -- -=---~; -- ~ , = --~ "- ~ --L~ ~~-~- ; 00 ~ - --- -.: __ --_ ---+!_-_--+t-=_-- --+l- --..L!- -.::J+: ----<»--+--'---~_- : ;---+1'.-- .-----:~-____:__:------:----- -l..! _---:- _ __ -r~~~_i ---ee ... ~ t-j:p_··=-~=~ =T:.:j: __ -.ll_- -+~--+I~=-J.: ---:J~ ~. _ ~L _ :=- DEPTH : 205 mm ~ . . ~ --.: II-":~-l---';~~=--l----:::~I~ ' ; ~- t- J r : i! - f____: , r-- . >---\ : , - - ! J , i :; .';: i ~; __ 1- - _ 1- i r I --t-o I- ~ ( l ( l ( ~~l ~--~ i ~~:~l---~11~6 4~B~~j~- __ j-~ -I __ ~~:~~~1§-~t-~- ~~~~~- ~~~-}~I~--~~-$-~!I~$~~~:~~t-~~I~*I~~;~~I ~~~~-!~=.~~-~~~~-=~-.~~~ i, i~i ;-1-'---1-lj ~I-~ -~ ~~ 121 ' ------4' ._L , -t-~=1=-=~=:J==1=~--=---=t~'-=t:--:t===~-=t=i==~f--=-=t=dr--::~-~ 1~ r-: 1- t- - - . - t- -- _ _1- =t=" ·1 T -- 1= I -- I: r --- ~ -I t---i- ~. . p; S dS ·ON ""(f5 ''' f ?fJ ( ( ( i. Il l c o 0 • S dS 'ON HllfH:J T'''(f5 ''''!?fJ 0 , S dS ·ON l~IfHJ ( l , , '- '- ( ( ( l ( c. II l Il II • • , 1 ! -~--t- - +-~ : 1 - ----t-- i -- r --r----r-. , ! , , - l-__ _ , l -~- ---+- I ----=---- ------- --- -- - - --- -- - ---(yr' ~-~--- _ _ _ __ "' _ _ o _ ____ ~_~ _ _ _ --c---------- ------ - - - 0 \ <; - dS 'ON UlliHJ ( ( ( --, :---t -~I -- --' --- ( SOIL DAlE LOAD DEPTII SENSITIVITY RUN TSPC. Series 2 9/1/75 111b 245 mm 5V 5 : ! l ( ( 1 ____ - ( , ,--- ( t -- ------ ----~----'--- __ l-_ _ -_ _____ __ __ __ =,~----~--------o-- -- ' __ 0 _ _ _ _ _ _ _ ____ _ _ - --' -- =----- ---"-- - -- 1== : -l. _ _ _ _ , -------i- _ ______ -r--'-, - -'-- ; ----~--, - : -----' -- ; ;--l - -j o o <;- t ( o ( ~, l I • • , , 1 ! I : i ! . i -~----;---t----i,~----.- . ----~-----.'-' -- ----'---'-, --:---_:..' --. , , , - - -N>-------------____ '--____ _ --.--_~ tv__ ------- -.-_. --~--b>_ .- . - :-------_._--------- SOIL DA1E LOAD LPC. Series 2 29/1/75 lllb -0V. 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M- c J , ----i!~(-·~· -i=~f~' ~:::~-=~~I-~t~.~~~'Et4--~~i~~----=L~~·:---·-.-j'~h·;=::·~~ -:~'~--:' ~-~:~ -.~ i~~i·,~--~";;- 9'F=j;;~IF~'~· ==~; ::ti , t~'~' ~ .~'~' '~~!';' ~f+' ,:.,,:' l=L~t~~·;Ft: ~-f:' f-= -=~i _~ _L' =1-i ~~L, ~~' -=~- __ "_ -=-~: ~~;! ~=' ~ .!-- - : . ---,' ---'.- . -~. 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' .::--. i 'I' : .. ~ _ . : - .----- - ._-+-++-..L--...:--t--,----,------'+: z:: -t~-------:--T--------- . - ·c---=..;:...-.:........~---=-----.-:.-----'----'-.....:.--j:--~+--------:~--+-T l---~++--t--~T-::------:.-----:-'-:---:=-::--. ------~-_____:~:. ---~.~-.. - '--",,---+. -f------.+---'-----"--.---.... ---.-----,--- 1 I' - -- .- - 1-_- -- - - 1- , k ,.- - :: - . - - - -r- -!-- DEPTII SENSITIVITY CHART SPEED : 255 mm 5V 10 mm/min 6 / . . --'-11- : __ ~ : I __ :: ______ :\_-:;s.".,.+_~~._'~.~-___ ·_---:;;;I_/~: -.-.. -- -' ._.=_-=.. _. ___ -____ ~ __ . ___ ~ __ :, __ .__ -_,, __ : __ -_--. __________ , ____ -__ :_~_,_ i - -'----+--.:..~'·-_11-.:---- ----~.--------:----------.-__ --=_:::.._ . : ____ ] __ : __ 1_---'-:_+-___ .-_-_-: __ .___ . ______ ~ _________ -_:_'_. _ tn ___ ~:_ .. _ ____ .· :..-l.-.:.:_f..:----.:..-·---'-----'-i----~--'-...!...t+-t-?fT-:-*/.--.-.-:------:__----.---~:-~:-_._.--,,~------~-------.-------'-: ---::~-;';r------'--~--'-~-'---'-__ --_ i _ _ . ______ . -== ---\ _. __ . ____ ~c--! - ~ .. - ______ .l. . __ ._._-;. . j - , ... --. --.j:o.--~ ......... - .' ._ -----\- -- .--~--- L I ·~IIr-_~_:: __ +-__ :: __ 1-_~_:'_:H~1 0,75 77 35 1,0 93 42 2,0 122 55 3,0 138 62 4,0 149 67 5,0 158 71 6,0 165 74 7,0 171 77 ._ _ 8,0 177 79 1I~-1-0,-0-+--1-85--+--8~3--,ll 12,0 189 .85 14,0 195 87 16,0 199 89 18,0 203 91 20,0 205 92 25,0 210 94 - - '--E::r1--O~- - _ _ . . _. ___ . --- --T ---1------ ----- '<--'=" -- .. - -.__ 30,0 214 96 IIr----+-----~----,1 35,0 217 97 - --~ ._--_._--"- ._--- _. 40,0 219 98 45,0 221 99 - --e')- . - ----~----.---_-. "--- -- .-.-------- .----CJ')--.-- ..... - 48,0 223 100 ( _. ., ( - .---{ - . J I' -"' I J -_ ! . r -=-= '.=--- -=r-=-=l=-==-t -- . =-~----~"-=:~ -I=' -=- -1--I---i---+-. i~---~'i-..-~-·-· .--- t.-- I L -+- + -i--1-' --,-_:.-=c=: ':::=1 :::r- .. ~ I ! i --r-- ,- - 0 '-1'- +- ~r ~- ,, - II • I '---·--;!--f----::-I-- - - , ~' I - -: i f-- , -t- _ _ _ . __ ~-+----.--....,i-r--+ j 1 - ! -. i - • -_~-+; -- ..,!r_------t----T" -~------~-.-: -~-~ -1 i ;-- I ---~'I - ~t= ~ --~f -=-~ ~ r ........L •. -= = ~ - .. ! i ~ -. I '=--:j' -i-- =:f -- ~. - -=+-=-~-- . - i I ! I· I j + ~;- -- - 'Ii : - L--.t==-; . -- . 'd I _ -i ~ t -- I,' -, i- -I- I - : -- I - r tv-+-...,.---_...l.I-, -+----+------.-.........,-- . i .' =r -r t- - , i t- 1=1_ -- Lt--! ~ I - t-::- ;-~ j f -=-=t 1 I ~: i~ =re i -' ........=--=--. ':::..=....t' I .=~::.....- 2-.~!=-:........=t:!=- =- ..;~=:=.;..:==+1=· =+=1 '=' -="+1 =' =--=-'==-1=+=: -==-:-=-==+=~=~_..:.......--....:-~--..L_+__----.--~--- _L_-.:.:::' ... +..~-::....:...::.l _~ _ -~--r--:- --t- --~. ---- , '" .11--- , '_' ,,: ' I--+! -=l--~- + .. _== __ +----- .. ~ -= ,_ _ , " -,_. _____ ' _ _ ·_- _·-.,. .... . -.:i,_· _ _ ·_~ ..,.-i·_- _l---r-- - : -=-~~~ .. -. ----1- ---- ~-: t- --=--~I:.·~~!~--- - =.i,--~=~-= - ~-'==r==-! -~-l -- - "-' ~-r' ~--. ·. I, - i .. ~ .,' i ___ . __ l' ! -1_ ---., , , \ -- ..j..~ , 6>-:----.-~-~---~,----------.- -I ,.., . .,.-- -. =.';- ~= =~- _____ . ~F~-- : ' -', : -~-.. +-- - ~,. ' ' ______ ~ - ---'. . --+I-- --....L..--- :-------i--------+-----,.------1:----~---- --,-"---"---'- - --.. "-~-. ; .. - ----.. -.".----- . ' 1, t, ' :---/ -.~ -. -=:..1'--....:-:. ·-t '-'--1- ---' -~,' ----'- ___ ~p--+t--------"·-----'-~----:_· .- -------.----- ,--- - ,,_ ,-1-1- - - --- ------+--:----+--- ---;-'.-----""""T". - ----+-- - - ........ -.- --- ,-----.. - ----------.- - --- -~ ~~; =-=-_-' -_ .=-:-_._ --::: 1 - . J ------+,----_ ..-;'--. -.----;====--. --==----~----.- .. - ----- -. ----- ".- .-.---.------~-------.. ---_._- --_. ! -,-~----- --.~ - ~....! r I ! ; .. -~--~--. . -'------- " .~.--. . -. I , --------------tu~.~-----~----------------------------·------u~~~I ---------------------~---------~r_--------~U~"---------------------,------- , . -t -=1 --=- -====-: t ---.-----.--~-----------------.~.-... ---.--- I - I - 1 , , - ---'-~-cr"'" - ,'--------- ---.---.-- ------- ---.--- .--'--~--------.-... c..,'-.-, --- ---------------.. ., ~--.-~ . - ___ 1 - _. - - - --.. ---~- .. ---.---------'--- ----- - - - --- { ( ( ( l ( ( ( ( ( ( 20,0 274 24,0 286 26,0 288 28,0 ' 288 ...... ,- ----- .'0:- APPENDIX III GRAPIDCAL EVALUATION OF SEITLEMENT RECORDS GRAPIDCAL EVALUATION OF SETTLEMENT OF RECORDS Extensive use has been made of Asaoaka's (1978) method to determine <;, from the settlement records. This is briefly described below and illustrated in Figures lll.1.a; rn.1.b; rn.2; Ill.3 and rn.4 1. The observed time-settlement curve plotted to an arithmetic scale is divided into equal time interva~ At (usually At is between 30 and 100 days). The settlement PI' P2 etc corresponding to Atl> At2 etc are read off and tabulated. (Fig Ill.1.a) 2. The settlement values PI' P2 etc are plotted against Pi-I' ie as points (PI' P2) etc in a coordinate system of Pi' and Pi-l' The 45° line Pi - Pi-I' is drawn (Fig Ill.l.b) 3. The points plotted in (2) are fitted by a straight line. Where the line intersects the 45° line gives the end of primary consolidation settlement. The slope of the line BI, is given by: 4. i) -5 2 In Bl c.,=-d __ 12 At The plotted line will become straight only after the end of construction and represent only consolidation. ii) Sometimes two lines can be fitted with the second representing secondary compression. iii) The method is also applicable for embankments where vertical drains have been used, in which case the dominant drainage length will be the effective drain spacing. The significant advantages of this method are that it is theoretically correct and it does not require any knowledge of the final settlement. As with any consolidation analysis the evaluation of <;. is dependent of the square of the drainage path length hence this length must be assessed with considerable care. The method clearly separates any primary consolidation and secondary compression and where the latter is significant, it can be deducted from the total settlement for evaluation of elm' Asaoka graphical analyses were performed for every embankment and two of these for the Bot River data at different stages of construction are shown as examples in Figures III.2, III.3 and IIl.4. It may be noted that a linear regression line is fitted through the data points. The slope of the line and the point of intersection with the base line (45°) can be calculated, or obtained graphically, to give BI , for use in the equation in (3) to give <;. and also to directly give the settlement, hence time, at the end of consolidation. In the examples the horizontal and vertical scales are not equal hence the so called 45° line is not at 45°; this is taken into account in the calculation of the slope B1• Time I t a) PARTfIlON OF SETTLEMENT RECORD INTO EQUAL TIME INTERVALS (1) Pen I I I I I ~ I I 0:- I ~ I I I P2 I I I I I I 0 ~ Pz ~ Po;) Pi-1 b) PWT OF SETTLEMENT VALUES AND FITTING OF STRAIGHT LINE STEPS FOR THE USE OF ASAOKA'S METHOD FIGURE No. ][ . 1 BOT RIVER STATION 8 + 830 Plate no.4R 8 2,~ 7 ,. -2,0 ~ r 6 ~ "" (~ _.-'- -,§. -' ,../' ~ 5 "'. ~ /' J r '''-.' -1,5 ~ h: 4 / ~ ~ ~ -,- ., ~ -' ~ -' r-I,O ~ -' ij 3 j ,,;' .- ....J I ~ ::::! 2 V r-0,5 ~ l( / / . I / • .r oJ 0 I I I I 0 0 200 400 600 800 1000 1200 No, OF DAYS SINCE 9/05/76 LEGEND Fi II Elevation. Plate Settlement. -'-'- FIGURE No BOT RIVER SETTLEMENT RECORDS ill . 2 500 450 400 350 300 ~ 250 ~ ........ '=. 200 '0 /50 /00 50 0 0 50 100 BOT RIVER STATION 8 + 830 Plate no. 4R (200-500days) /50 200 250 300 t (i-/) mm ASAOKA PLOT 350 400 450 500 FIGURE No ]I . 3 ~ ~ BOT RIVER STATION 8 + 830 Plate no.4R (700 -900 days) ~ _A 200 1 ~ / ~ :7 150 I :;/" ~ '0 100 I :// 50 I 7/ o r--= I I I I o 50 100 150 2CXJ 250 t (i- f) mm ASAOKA PLOT FIGURE No ][ . 4